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November 9 – 10, 2006 University of Coimbra Coimbra, Portugal
Dear Reader We, as the Management Committee of COST Action E29, would like to extend our warm welcome to you, the reader of the proceedings of this 3rd International Workshop on Earthquake Engineering on Timber Structures, held on 9-10 November 2006 in Coimbra (Portugal). This event was organised by: ¾ Department of Civil Engineering, Faculty of Sciences and Technology,
¾ Department of Civil Engineering, School of Technology, POLYTECHNIC
Earthquake Engineering is one of the most important safety subjects in the performance of buildings, elements and components. Earthquake Engineering is a specialised field which plays an important safety role in our everyday lives (in general). Earthquake Engineering in lightweight structures such as timber frame buildings plays an important role in safety of people in and around buildings in the World at large. The proceedings of this conference therefore make an important contribution towards the enhancement of the understanding of earthquake engineering in lightweight timber and timbercomposite buildings. It is my pleasure, as the Chairman of this COST Action, to have witnessed a series of excellent papers produced and presented in this Workshop by some world experts in the field of Earthquake Engineering for timber structures. The Management Committee and I would like to extend our thanks to all the delegates of the conference, authors and presenters without whom this conference could not have been such a success.
Dr Vahik Enjily (Chairman of COST E29)
Organizing Committee
COST - European Cooperation in the Field of Scientific and Technical Research Action E29: "Innovative Timber & Composite Elements/Components for Buildings"
UNIVERSITY OF COIMBRA Department of Civil Engineering Faculty of Sciences and Technology
Department of Civil Engineering School of Technology
Sponsorship .
17 "Rehabilitation of Lisbon's old "seismic resistant" timber framed buildings using innovative techniques" Raquel Paula and Vitor Cóias …………………………………………………... 71 .... Paulo Cruz.. 1 "Selected research and case studies in ancient Portuguese structures" Paulo Lourenço and Ricardo Brites ……………………………………………. Mauricio Piazza and Humberto Varum ……………. 33 2nd session: "Design of Timber Structures" "Utopia pavillon project" Jean Paul Perrin ………………………………………………………………..Application for timber structures" Eric Fournely and Thierry Lamadon ………………………………………….Table of Contents 1st session: "Traditional Construction" "Dynamic modelling of timber joints in traditional structures" Jorge Branco. 47 "Design guidance on Eurocode 8 for practicing engineers for timber structures" Tomi Toratti …………………………………………………………………… 55 "Design detailing for earthquake engineering .
Kiril Gramatikov and Roko Žarnić "Pseudo-dynamic test method.3rd session: "XLAM Innovative Wooden Structures" "Seismic Behaviour of Multi-Storey XLAM Buildings" Ario Ceccotti and Maurizio Follesa …………………………………………… 81 "Study of Lateral Resistance of massive X-LAM wooden wall system subjected to horizontal loads" Bruno Dujič and Roko Žarnić .......... Ágnes Ruff....... 135 5th session: "Joints" "Seismic Performance of Timber Shear walls sheathed with OSB panels" L............... Bruno Dujič..... 175 .......... Fülöp... 105 4th session: "Shear Walls" "Quasi-Static and Pseudo-Dynamic Tests on XLAM Walls and Buildings" Marco Pio Lauriola and Carmen Sandhaas ……………………………………..... 163 "Dynamic performance of wood based panels in roof and floor construction" Mizi Fan and Vahik Enjily …………………………………………………….... Dubina ……………………. 97 "Influence of openings on shear capacity of massive X-LAM wooden walls" Bruno Dujič...A.......... Bálint-Major and D. 119 "Influence of the loading protocols on the Hysteresis response of sheathing to framing nailed joints in shear walls" Tatjana Kočetov Mišulić....... S. experiments and accuracy" Kjell Malo ……………………………………………………………………… 143 …... Simona Klobcar and Roko Žarnić ………………………………...........
Each solution has unique consequences in terms of the joint final strength. even without any strengthening device. the need arises for developing behavioural models of these joints.it Paulo CRUZ Associate Professor University of Minho pcruz@civil. The study addresses plain timber connections. but realistic.piazza@ing. These devices are extensively used in structural upgrading operations. usually have a significant moment capacity. particularly for steel and composite structures. Common constraint models. but should be using semi-rigid and friction based models.ua. indeed. Experimental results show that structural response of traditional timber connections under cyclic loading cannot be represented by common constraint models. A research program has investigated the behaviour of old timber joints and examined strengthening criteria. 2. From this consideration. Little attention has been devoted to the joints in old. Joints strengthening can be done in a number of possible ways: from simple replacement or addition of fasteners. cannot satisfactorily describe the real behaviour of these joints.uminho. considerable research efforts have been devoted to characterise the semi-rigid connections behaviour. traditional timber joints. several studies have covered the problem of modelling the behaviour of new engineered connections.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. The work presented here has been developed within a general research program devoted to the definition of synthetic models for the static and dynamic behaviour description for common timber connections in traditional.unitn. to be used for the investigation of structural dynamic response [1]. like perfect hinges or rigid joints. A synthetic model of cyclic behaviour has been adapted on the basis of experimental results.uminho. timber roof structures. 2006 Modelling of timber joints in traditional structures Jorge BRANCO PhD student University of Minho jbranco@civil. connections that are strengthened by steel elements. for example. as well as. Keywords: Cyclic behaviour. In Portugal and Italy. to the use of metal plates.pt Humberto VARUM Assistant Professor University of Aveiro hvarum@civil. The joint behaviour may be classified as semi-rigid and. old and non-engineered constructions. models for the joints in old traditional timber structures generally leads to very conservative retrofits and upgrades to satisfy new safety and serviceability requirements. in order to develop a reliable response in the case of cyclic loading. experimental testing 1. timber are generally impaired by the inadequacy of commercial finite element software in modelling the partial restraint to rotation and the limited moment transmitting capabilities of their connections. The main parameters affecting the mechanical behaviour of the connection have been singled out. strengthening. Portugal November. When structural analyses have to be carried out for evaluating the possible need and effect of strengthening. Introduction In the field of timber structures.pt Summary Original unstrengthened timber connections and the effects of different strengthening techniques have been evaluated experimentally with tests on full-scale birdsmouth joints. are part of the constructional tradition. is influenced by the time-varying level of compression between the joined members. like hinges or full restraint connections. in particular. glued composites or even full injection with fluid adhesives. The lack of practical. stiffness and ductility. traditional structures that are very frequent in Europe. being based on friction. For the case of steel 1 . Model behaviour of semi-rigid timber connections In recent years. also in earthquake prone areas. Traditional timber joints.pt Maurizio PIAZZA Professor University of Trento maurizio.
[3] derived a hysteretic model for moment resisting nailed timber joints. Kivell et al. In some typical structural configurations of timber constructions. Semi-rigid modelling of steel connections in seismic design acknowledges and exploits the dissipation capabilities of their hysteretic behaviour.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. In the case of existing structures to be rehabilitated or eventually upgraded. The semi-rigid modelling of timber connections. For deformations larger than those that have occurred already in the joint. Dolan [4] and Stewart [5] each developed hysteretic models to describe the cyclic behaviour of timber frame walls with nailed connections between the framing and the sheathing. A numerical model for these connections must. Although quite different in detail. the commonly used hinge models are inadequate. As recognised by Eurocode 8 [2]. Ceccotti and Vignoli [6] modelled the pinching hysteretic for moment resisting semi-rigid timber joints made from glulam and drift pins. be sufficiently accurate to describe properly the semi-rigid behaviour. and sufficiently simple. a realistic interpretation of the global structural behaviour is a primary need. the equilibrium conditions may not be reached analytically. (a) Kivell et al. Different hysteretic models for timber structures have been developed (Figure 1). to allow use in common practice. with appropriate values to be obtained either by predictive models or from direct experimentation. the case of partial flexibility has been early recognized in design codes. both conceptually and computationally. according to new requirements. all the models follow the envelope or skeleton curve of the connection describing its behaviour under static loading. the seismic design acknowledges and exploits the dissipation capabilities of their hysteretic behaviour. 1 Hysteretic models for timber structures (d) Ceccotti and Vignoli [6] 2 . intends to represent the seismic behaviour of timber structures with a comparable level of detail for all the structural components. using nonlinear moment-rotation laws and hysteretic rules. where joints have moment resisting capability. because in real structures. all these models clearly describe the degrading stiffness for repeated load cycles caused by the plastic deformations of the wood surrounding the fastener. 2006 structures. a similar approach is appropriate also for traditional timber elements and structures. [3] (b) Dolan [4] (c) Stewart [5] Fig. With these models. Portugal November. then.
The axial force. there seems to be no shortage of hysteretic models that can be used in dynamic analysis of timber structures. The three basic types of intervention considered in this study are modern implementations of traditional strengthening techniques: the stirrups. The first step of the loading procedures in both the monotonic and cyclic tests was the 45 220 application of an axial compression force on the 75º rafter. the elements have a cross section of 80 x 220 mm2. are: 1) selecting the appropriate hysteretic model that should be incorporated in the global structural model and. a transversal millimeters) force. 2 Connections geometry (dimensions in the subsequent loading steps. The main challenges.02. it is defined as the negative direction. 80 Tests were performed under displacement 30º 422 control. a series of tests on unstrengthened A specimens were performed in order to 02.03 LVDTs _ _ in Figure 3. Metal devices were developed and applied basically intending to counteract out-of-plane actions. Nowadays. when the development of industrial production methods made bolts. N Channels Transducers including load cells and linear voltage 10 differential transducers (LVDT). 2) developing system identification procedures [7].01. F. and includes monotonic and cyclic tests of full-scale birdsmouth joints skew angle of 30º [8]. this practice became common only in the 19th century.11 Load cells 11 F + F d d 00 Firstly. Type and location of instrumental channels. 3 . and is intended to avoid the detachment of the connected members. as represented in Figure 2. are shown 00. Together with those that have been proposed for timber joints and structural systems. 3 Testing apparatus and instrumentation layout ancient times. 2006 There is a wide variety of available hysteretical models in general structural dynamics literature. then. rivets. Experimental tests campaign of full-scale timber connections The experimental research was carried out at the Laboratory of Structures of the University of Minho (Portugal). was kept constant during the test. simulating the effect of the self-weight and dead load presented in the 80 structure. However. the notch depth is 220 45 mm and the notch length is 422 mm. In Fig. and other metal elements easily available. which could not be resisted by the assemblage itself.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. the internal bolt and the binding strip (Figure 4). + 10. and particularly for the traditional timber connections. a set of joints were strengthened with metal devices and tested 01 under monotonic and cyclic loading. it is defined as the positive direction and when the skew angle decreases. Portugal November. Metal connectors have been applied occasionally in timber joints since very Fig.03 Section A-A characterize the original behaviour of joints 01 representative of existing timber systems. A 02 03 Subsequently. acts perpendicular to the rafter axis. strengthening also concerns the behaviour of the friction-based connection in the plane of the structure. When the skew angle increases. 3. For all the specimens.
Portugal November. The monotonic tests were carried out both on unstrengthened joints and on joints strengthened with basic types of metal connectors. BS3 BS4.5 1. 4 . A6 A7. represents. This pseudo-plastic phase. BS6 BS7. A permanent compression force of 25 kN (corresponding to 1. The experimental analysis can be extended by numerical models in the next research step.4 1. as well as.4 and 2.4 and 2.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.4 1. meant to inquire the properties of the connection in terms of initial strength. experimentation gave an insight of the joint behaviour for the calibration of the models. It was particularly important to observe the post-elastic behaviour and the failure mode of the connections for each situation analysed. for common Portuguese timber roof structures and according to National standard [10]. The tested specimens could not cover all the possible ranges and combination of parameters (as geometry. Table 1 summarises the test campaign conducted on birdsmouth joints (monotonic and cyclic tests). and the second jack imposed a monotonic transversal force (Figure 3). compression level in the rafter.4 1. 2006 (a) Stirrup (S) (b) Bolt (B) (c) Binding Strip (BS) Fig. B5. A8.4 1. A5. loading test velocity. Monotonic tests. after which became non-linear but only within a small range. a quasi-perfect plastic behaviour appears. B6 B7. S2.5 1. the dead load applied and the stresses derived from the serviceability limit state. The test results illustrate perfect elasto-plastic behaviour for the three curves (Figure 5). starting at a 10 mm displacement. B2. and post-elastic deformability. The first set of connections tested was composed by three unstrengthened joints (A1.4 The compression stress levels in the rafter adopted: 1. BS5. A2 and A3).5 MPa. S9 B1. The behaviour is perfectly elastic until the elastic limit displacement (≈8 mm).4 1.4 MPa and 2.5 1. A3 A4. B3 B4. A9 S1. S8.) that are of practical interest. B8. stiffness. A2. remains practically constant until the maximum displacement (around 50 mm). Beyond this. respectively [11]. etc. to allow the calibration of numerical models. S5. BS9 Type of connection Unstrengthened Unstrengthened Unstrengthened Stirrup Stirrup Stirrup Bolt Bolt Bolt Binding Strip Binding Strip Binding Strip Loading method Monotonic + Monotonic Cyclic Monotonic + Monotonic Cyclic Monotonic + Monotonic Cyclic Monotonic + Monotonic Cyclic Rafter compression stress level (MPa) 1. B9 BS1.4 1. BS8. S6 S7.4 1.4 and 2. BS2.4 MPa compression stress) was applied to the rafter throughout the vertical jack. 3.4 1. 4 Traditional strengthening techniques evaluated Tests on assembled connections were preceded by accurate material characterization. presenting a small decrease of the resistance after 25 mm displacement. S3 S4. Subsequently.1 Monotonic tests Test data of original connections have been gathered with the purpose of characterizing their behaviour. Table 1 Tests on birdsmouth joints Specimen A1. in terms of the mechanical properties of the timber elements used for all full-scale models [9].
The curves. with the decrease of the skew angle (negative direction). according to the scheme in Figure 3. for what concerns the initial stiffness characteristics. the brittle behaviour is substituted by a pseudoplastic phase. is observed in both cases.5 MPa Displacement (mm) Fig.4 MPa 2. a more brittle behaviour was detected when the skew angle decreases. However. 5 . followed by a loss of friction. a higher initial stiffness is also achieved with the increase of the compression stress level in the rafter (Figure 6). after which a slip. This ductile behaviour is due to the local compression of wood. the positive field. Portugal November.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. apart of an increasing in the maximum strength. 2006 10 9 8 7 Force (kN) 6 5 4 3 2 1 0 0 5 10 15 20 25 30 A1 A2 A3 35 40 45 50 1. remain nearly constant. a total loss of friction occurs with the failure of the connection. for the monotonic tests conducted for the two compression stress levels.5 MPa A4 A5 A6 Fig. Brittle behaviour. after the achievement of the elastic displacement limit. only an increasing in the maximum force and corresponding elastic limit displacement can be observed. 6 Force-displacement curves obtained in the negative field for two rafter compression stress levels Comparing the force-displacement curves obtained from the two different compression stresses in the rafter levels. 5 Force-displacement curves obtained in the positive direction for two rafter compression stress levels For the opposite direction of loading. induces a rapid decrease of the resistance. After the new stable position of the joint is reached. Finally. average values for the 3 specimens. the behaviour of the curves is similar. The curves presented in Figure 6 show a behaviour perfectly elastic just at the maximum force. Table 2 summarises the main results.4 MPa 2. Displacement (mm) -24 -22 -20 -18 -16 -14 -12 -10 -8 -6 -4 -2 0 0 -2 -4 -6 -8 -10 -12 -14 -16 Force (kN) 1. In the other direction.
72 -10. it is recognized that all the strengthening schemes analysed increase the stiffness. 15 10 5 Force (kN) 0 -5 -10 -15 -20 -25 -50 -40 -30 -20 -10 0 10 Unstrengthened + Unstrengthened Binding Strip + Binding Strip Stirrup + Stirrup Bolt + Bolt - 20 30 40 50 60 Displacement (mm) Fig.32 Stiffness (kN/mm) x 103 Fe50% / de50% Regressio Fe / de n 674 634 647 1771 1785 1958 1569 1389 1408 1705 1661 1758 Comparing the test results in terms of force-displacement curves for the unstrengthened and strengthened connections (Figure 7). In particular. however. Portugal November. in the positive direction and the maximum resistance for both directions. 2006 Table 2 Influence of the compression stress level in the rafter in the response.4 2. in the strengthened connections with binding strip. when the tests finished. the maximum resistance for the strengthened connections with stirrups and internal bolt is achieved near the end of the test. the first one is more efficient in terms of ductility capacity with the goal to assure a better seismic behaviour of the joints. between the internal bolt and the binding strip.75 10.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Connections strengthened with stirrups and binding strip attained the same range of maximum force.84 -15. the less efficient.31 5. Only the binding strip showed limitations in terms of maximum displacement. while the elastic stiffness are similar. However. 7 Force-displacement diagrams for original unstrengthened and strengthened connections under monotonic loading 6 . However. Comparing the strengthening techniques evaluated. The benefits in terms of stiffness are not so significant.76 5. for monotonic loading. in particular. of the joint σc (MPa) 1. is the internal bolt. the brittle behaviour exhibited by the original unstrengthened connections disappears in the strengthened specimens. the main profit of adding a metal device to the joints is ductility improvement with clear advantages in the seismic response of these structural connections. the force value was already decreased. The elasto-plastic behaviour with limited ductility evidenced by the unstrengthened connections is substituted by full non linear curves exhibiting high ductility in the strengthened connections.5 Direction (+) (-) (+) (-) de (mm) 8. The effect of the strengthening schemes in the negative directions for the monotonic tests is quite obvious: the maximum resistance and the ductility capacity increase. in terms of maximum resistance.47 8. Therefore. this last scheme has a lower ductility capacity. Therefore.13 Fmax (kN) 6.
0 de σc d d + F+ F - 0. 1. as well as. for both positive (de+) and negative (de-) directions. This sequence is in accordance with the proposal in the recommendations of the EN 12512: 2001 [12] (Figure 8). one cycle in the range [0. 3. 9 Influence of the rafter compression stress level in the cyclic response Observing the cyclic response of strengthened connections and comparing with the original 7 .25 de+.0 de Time Fig. 15 10 5 Force (kN) 0 -5 -10 -15 -20 -20 -15 -10 -5 0 5 10 15 σc = 1.25 de-]. 0. came directly from the results previously achieved in the monotonic tests. The values used for the elastic limit displacements. Portugal November. The energy dissipation grows with the compression stress level in the rafter (2.50 de-].5% to 3. the test program included one complete cycle with the amplitude [0. were imposed. 0. The first observation taken from the cyclic tests on unstrengthened connections is the different response for each direction of loading. ….4 MPa σc = 2. with increasing amplitude.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.50 de+. the force-displacement curves presents an increment in the maximum strength (Figure 9). Significant energy dissipation occurs only in the negative direction caused by the sliding of the rafter when pulled into this direction. In particular. Veq).75 de+.75de 1. three cycles in the range [0. Increasing the compression stress level at the rafter. (1+n) de-] with n = 0.0 de Displacement 2. until the failure of the joint. three cycles in the range [(1+n) de+.75 de-].2 Cyclic tests In these tests a reduced number of cycles.5 MPa A9 Displacement (mm) Fig.96% in terms of hysteretic equivalent viscous damping ratio. 2.0 de 4.0 de 5. connections strengthened by the techniques presented above in Figure 4. 8 Cyclic loading procedure adopted Cyclic tests have been performed on various series of samples of plain timber connections. 2006 3. 0.
17 -39.38 13.57 dmax+ (mm) 16. the energy is dissipation increases. Without strengthening.5 MPa) Binding Strip (BS) Bolt (B) Stirrup (S) Dissipated Energy (kJ) 230 380 2874 1877 1859 Veq (%) 2. on the strengthened and original unstrengthened joints with a rafter compression stress level of 1. The number of cycles achieved is particular important taken into account the Eurocode 8 [2] imposition for the behaviour factor.75 Fmax+ (kN) 6. Portugal November. and at a static ductility ratio of 6.20 9. 10 Force-displacement response for the cyclic loading: original unstrengthened and strengthened connections The experimental force-displacement diagrams achieved for all connections are asymmetric.29 18. both in terms of stiffness and yielding strength. can be concluded about the positive and negative effects of the strengthening techniques studied.30 -21.38 15.30 28.49 9.96 6.63 -35. 2006 unstrengthened.00 -25.85 11.47 -21. the joint is not able to prevent the failure caused by load reversals (detachment of the connected elements) and therefore the energy dissipation capacity is very low.68 dmax(mm) -15. it is referred that the dissipative zones shall be able to deform plastically for at least three fully reversed cycles at a static ductility ratio of 4. under cyclic loading. The hysteretic equivalent viscous damping ratio (Veq) evaluated from test results is considerable (Table 3).4 MPa.45 3. All strengthening techniques adopted were efficient in the improvement of the cyclic hysteretic behaviour of the connections.45 23. In this standard.09 Fmax(kN) -11. 20 Force (kN) 10 0 -10 -20 -30 -40 -30 -20 -10 0 10 20 30 -40 -30 -20 -10 0 10 20 30 Unstrengthened Stirrup 20 Force (kN) Force (kN) 10 0 -10 -20 -30 Displacement (mm) 20 Force (kN) 10 0 -10 -20 -30 -40 -30 -20 -10 0 10 20 30 -40 -30 Binding Strip Displacement (mm) 20 10 0 Bolt -10 -20 -20 -10 0 10 20 30 -30 Displacement (mm) Displacement (mm) Fig. Figure 10 collects the force-displacement diagrams.4 MPa) Unstrengthened (σc=2.15 18.57 -17.83 -21. 8 . Table 3 Main results for the cyclic tests on the original and strengthened joints (average values) Joint Unstrengthened (σc=1. under cyclic loading.60 With the increase of cyclic displacement amplitude. without more than 20% reduction of their resistance.28 14. for ductility class H structures.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.08 -15. for ductility class M structures.
1 Model parameters When modelling experimental test results with a synthetic law. after an excursion in the non elastic field. φy+.4÷0. 9 . My+. the relationship between the axial compression level and the limit moments needs to be expressed and included in the synthetic model. One characterising aspect of the friction connections under study is that the ultimate and yield moments. increasing skew angle) or k0. for the positive. Consequently. − the maximum moment and maximum rotation at first loading.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.9)My-.8÷0. and kP-. φy-. Indications for assessing their values and those of the parameters in the following when the non elastic behaviour field can be easily identified or when the mechanical behaviour of the joint exhibits a continuous change of curvature are given in [12].9)My+. or at the value My+ . kP+. k0+ (positive field.90÷0. In order to define it. determined on the curve of monotonic behaviour. for positive or for negative rotation (Mu+. φu-). − the moment and rotation values corresponding to the yielding (transition between the elastic and non-elastic behaviour). 2006 4.5)Mu-. or viceversa. after an excursion in the non elastic field. The former describes completely the joint behaviour for structures under monotonic loading. M-φ. or (0. the latter requires defining additional values. the characteristic intersection points between branches must be derived from the experimental moment-rotation diagrams. or (0. φP1+) and of its analogous P1. for the two cases above. significant variations of axial forces in these elements may be expected to occur during the strong motion phase already at medium intensities. characterising the transition between the elastic and non elastic behaviour. a different level of approximation may be adopted for the first loading curve and for the cyclic case [13]. The adoption of a hyperbola permits to describe quite precisely the transition between the elastic and the plastic branches. Portugal November. instead of the two values kR+ and kRThese last parameters may be easily identified in an M-φ diagram. − the residual stiffness. determined on the straight line connecting the origin and a point of the monotonic curve corresponding to moment values in the range (0. φu+. but a bilinear approximation may also be generally satisfactory. Mu-. kR. k1+ or k1-. also in the two cases above. it seemed reasonable to consider a series of cycles presenting maximum values of imposed rotation similar to those expected in the numerical application. or in the range (0. the following parameters have special importance: − the initial tangent stiffness. ku+ or kuThese first parameters are common to a monotonic and a cyclic model.8÷0. average value. they could as well be derived from monotonic ones. some simplifications in the mathematical description are advisable. Since the values decay. it is generally reasonable to assume a single.in the negative field. My-. For a cyclic model to be used in dynamic analyses. as a function of the extension of non elastic excursion. − the coordinates of a point P1+ (MP1+. − the average residual stiffness passing from the positive to the negative moment. In seismic conditions. M-φ. In a first approximation and in the absence of specific cyclic testing. To this purpose. respectively. for the negative quadrant. Numerical modelling 4. Further. Mu and My.95) My+. the point P1+ may be recovered at a value of the moment MP1+= (0. it has been proposed to derive them from the last cycle of a series of 3 at constant amplitude [14]. are directly related to the level of compression in the rafter. and a multilinear rule is considered generally adequate [1].4÷0. as follows: − the secant stiffness at complete unloading..5)Mu+. − the initial secant stiffness.
6 (MPa) Mu- Mu+ Fig. when is difficult to separate the elastic branch from the plastic one 10 . i.2 1. For structures under static monotonic loading. namely: a) when a clear distinction exists between the elastic and plastic branches.4 2.and the second branch by ku+ or ku-. and. respectively. The EN 12512:2001 [12] procedure suggests two different methods for the definition of the bilinear approach. depending on the degree of approximation required. elastic branch is characterized by a slope k1+ or k1. 2006 (kNm) 11 10 9 8 7 6 5 4 3 2 1. this first loading function describes completely the joint behaviour. while some simplifications may be advisable in the mathematical description. two different approaches are possible.2 Modelling of the monotonic tests The first step in developing a synthetic law of behaviour for the joint considered is to identify a law for first loading. Mu+ ku+=1/6 k0+ atn ku+ My+ 0. My+/ k1+) or (My-. and mainly in the application. for positive or negative rotations. As mentioned above.e. When considering the bilinear approximation. for the "basic curve". Portugal November. b) when is difficult to separate the elastic branch from the plastic one (Figure 12).International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. which may be particularly suitable to dynamic analyses from the computational point of view. in the cases of positive and negative rotation. 11 Ultimate moment Mu versus compression stress level in the rafter (σc) 4.4 Mu+ atn k0+ φy+ φu+ Fig. The transition point has coordinates (My+. the first. An accurate matching of the experimental curve and the model may be feasible by higher order curves.6 1.8 2 2. of a cyclic model.2 2. My-/ k1-).4 1. The method to use depends on the development of the experimental curve. 12 Method proposed by EN 12512:2001 for the bi-linear approach.
4. The third method applied supports the bilinear approximation which minimize the difference between the energy dissipated in the test results and in the bi-linear curve model (Figure 13).3 Modelling of the cyclic tests It is possible to follow the evolution of the behaviour of a connection under cyclic loading and to define the individual branches composing the corresponding linearised M-φ diagram.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Portugal November. another method was used in this work for the development of the bilinear approximation. was adopted in the numerical analysis in this work. The model used in this work is based in the work present by Parisi & Piazza [1]. However. and since the capability to dissipate energy is crucial in the seismic response. 2006 However. some corrections are introduced and the parameters are adapted based on the test results obtained in the experimental campaign reported in section 3. 11 . the third method. Because of that. using these methods. M Experimental curve Bi-linear curve model φ Fig. the synthetic law can present different amounts of energy dissipation when comparing with the experimental curves. which gives the same energy dissipation. 13 Bi-linear curve obtained from the minimization of the difference between the energy dissipated Table 4 presents the comparison between the experimental curve and the synthetic law obtained using the three methods in terms of energy dissipation. Table 4 Comparison between the energy under the experimental curve and the synthetic law given by different methods (kJ) Specimen B1 B2 B3 B4 B5 B6 Experimental 496 688 815 885 715 737 EN 12512 Method (a) 515 705 854 908 728 743 EN 12512 Method (b) 620 710 630 911 915 844 Bi-linear (Same energy ) 494 687 810 885 715 738 Tri-linear (Same energy ) 496 684 814 886 716 738 With the purpose to obtain a more accurate model.
thus. according to experimental evidence.+ (φ . My+/ k1+) or (My-. with the slope kU+ or kU-. is elastic. Branches (5) and (6) in the figure are calibrated for unloading.= My. For increasing moment beyond the elastic limit. My-/ k1-). M My + (2) atn kU (1) atn k1 + + atn k1 atn kU (4) - - φ - (3) My Fig.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. the second branch equation is given by: M+ = My+ + (φ . the slope of the plastic branch is not affected and. 14 Model for the cyclic behaviour.φy+)×kU+ (1) and M.φy-)×kU(2) for positive and negative rotations. part 2 (unloading and reloading) 12 . Portugal November. by a slope k1+ or k1-. respectively. The first branch. 2006 Figure 14 highlights the behaviour of the connection at first loading. part 1 (monotonic loading) Figure 15 describes the situation of unloading for rotations in the branch (2) or (4) and subsequent unloading. The transition point has coordinates (My+. for positive (2) or negative rotations (4). Moments may undergo variations depending on the level of axial force in the rafter. for positive rotation (1) and negative rotation (3). and characterised. for moments inferior then My+ or My+. respectively. but. the connection responds following a second lines branch. 15 Model for the cyclic behaviour. may be kept constant in the model. M My + (7) atn kP (6) - atn kR + (5) atn kP + atn kR - (8) My φ Fig.
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
The unloading stiffness is generally 50 to 80% lower than the elastic one, according to experimental observation [14]. A reduced stiffness, kP, is adopted for these branches, expressed by the equations M+ = Mmax+ + (φ – φmax+)×kP+ (3) and M- = Mmin- + (φ - φmin-)×kP(4) Paths (7) and (8) are followed after complete unloading from the negative and positive quadrant with change of sign of the moment, according to equations M+ = (φ (6) - φ)× kR+ (5) and M- = (φ - φ (5))×kR(6) where φ (6) and φ (5) are the rotations on branches (6) and (5) for null moment, respectively. These low-stiffness branches following unloading are typical of mechanical connections in timber structures. The corresponding residual stiffness, kR, may be very modest: experimentally, they amounted to 10÷50% of those relevant to unloading branches (5) and (6), depending on the type of strengthening technique. When unloading occurs from a limited plastic excursion, and because the stiffness kP is lower than the initial stiffness k1, branch (5) or (6) may intercept the first loading branch (1) or (3) before the null-moment axis. In this case, that is characterised by an intersection point with all positive (or negative) coordinates, unloading continues along the direction of the first loading branch. A reloading path along segments (5) and (6) will follow branches (2) and (4) upon intersecting them. In order to simulate pinching correctly, branch (7) continues in the positive quadrant until meeting branch (5) or (1), if previous cycles have reached, or not, respectively, the plastic range in the same quadrant. Similarly, in the negative quadrant branch (8) will join branch (6) or (3). From the intersection onward, stiffness kP or k1 is regained. This situation is shown in Figure 16, where subsequent loops activate branches (5) and (6) in loading and unloading.
M My (7) (9) atn kP (6) (4)
(5) atn kP
Fig. 16 Model for the cyclic behaviour, part 3 (pinching effect) After a first loop, linearization may induce, as represented in Figure 17, a slight translation of branches (7) and (8) toward greater moments, in absolute value. This effect is nonsignificant in terms of global response, also considering that, after a small number of initial loops, the connection will remain in the region enclosed by branches (7), (8), (9), and (10), as shown in Figure 17.
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
M My atn kD (10) (8) My
(7) atn kD (9)
Fig. 17 Model for the cyclic behaviour, part 4 (translation) Branches (9) and (10) are followed, in loading or unloading, upon moment reversal on the joint from behaviour fields (7) or (8). They represent the situation typical of a joint subjected to low intensity actions after being exposed to rotations of significant value. For these branches, stiffness values included between the initial elastic and the unloading ones are suitable. The same value may be reasonably adopted in both quadrants. The corresponding equations are: M+ = (φ - φ7-9)×KD+ + M7-9 (7) and M- = (φ - φ8-10)×KD- + M8-10 (8) where (φ7-9 ,M7-9) and (φ8-10 ,M8-10) are the coordinates of the points in branches (7) or (8) where the new branch takes over. From numerical analysis in similar connections [14], the influence of these stiffness values on the global dynamic behaviour is modest. In seismic conditions, however, after the strong motion phase, this area with low residual stiffness may still contribute with a limited energy dissipation. The physical experimentation has shown that, in these cases, it is the friction between coupled surfaces of the elements converging in the node that supplies this energy dissipation contribution in the structural response.
5. Conclusions and final comments
The typical birdsmouth joints, even without any strengthening device, usually have a significant moment-resisting capacity. Therefore, they cannot be represented by common constraint models, like perfect hinges, but should be considered semi-rigid and friction based. The test results show that this capacity is function of the rafter compression stress level. Moreover, it is clear that the width of the rafter, the friction angle, and the skew angle in the connection are also important. The experimental analysis has been of fundamental importance in order to understand the real behaviour, by pointing out some important aspects like force transmission mechanisms, failure modes and guidance for appropriate strengthening solutions. Strengthening, usually performed by insertion of metal devices, is indispensable for ensuring adequate joint response, in particular, for seismic loading, or in other adverse and unpredictable loading conditions. The strengthening of the joints results in a significant increase of the hysteretic equivalent viscous damping ratio (Veq). The energy dissipation became significant. In conclusion, the strengthening solutions studied improve the seismic behaviour of the birdsmouth joints typically presented in traditional timber roofs.
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
The possibility of modelling these connections numerically, by means of nonlinear moment-rotation laws and hysteretic rules, intends to represent the seismic response of timber structures with a comparable level of detail for all the structural components. The paper presents a model for interpreting the cyclic, semi-rigid behaviour of traditional joints in timber structures, along the guidelines of the European Codes. The model is based in previous works [1] and on the results of experimental results of full-scale birdsmouth joints. Next step of the research program will be the implementation of the hysteretic models in a finite element format.
6. Acknowledgments
The first author gratefully acknowledges the Portuguese Foundation for Science and Technology, for his PhD grant, with the reference SFRH/BD/18515/2004. The research described in this paper was conducted with financial support of the Portuguese Foundation for Science and Technology (contract reference POCI/ECM/56552/2004). This work has been carried out with a partial financial contribution of the Italian Earthquake Engineering Laboratory Network (RELUIS).
7. References
[1] [2] [3] [4] [5] [6] [7] [8] [9] [10] [11] [12] [13] [14] Parisi, M.A. and Piazza, M., "Seismic behaviour and modelling of traditional timber roof structures", Proceedings, 11th European Conference on Earthquake Engineering, Paris, 1998. prEN 1998-1:2004, Eurocode 8: Design of structures for earthquake resistance. Part 1: General Rules, Seismic Actions and Rules for Buildings, CEN, European Committee for Standardization. Brussels, Belgium. Kivell, B. T., Moss, P. J. and Carr, A. J., "Hysteretic modelling of moment resisting nailed timber joints", Bull. New Zeal. Nat. Soc. Earthquake Engineering, 14(4)(1981 ) 233-245. Stewart, W. G., The seismic design of plywood sheathed shear walls, Ph.D. thesis, University of Canterbury, Christchurch, New Zealand, 1987. Dolan, J. D., The dynamic response of timber shear walls, Ph.D. thesis, Department of Civil Engineering, University of British Columbia, Vancouver, BC, Canada, 1989. Ceccotti, A. and Vignoli, A., "Engineered timber structures: An evaluation of their seismic behaviour", in Proceedings, 1990 International Timber Engineering Conference, Tokyo, Japan, pp. 946-953. Foliente Greg C., "Modeling and Analysis of Timber Structures Under Seismic Loads: State of the Art", Earthquake Performance and Safety of Timber Structures, Ed. By Greg C. Foliente. Forest Products Society, Madison, 1997, ISBN 0935018859, USA, p. 5573. Branco J., Cruz P. and Varum H., Experimental Analysis of Birdsmouth Joints, Report E11/05. DECivil, University of Minho, 2005, 49 pp. Branco J., Varum H. and Cruz P., "Structural grades of timber by bending and compression tests", Advanced Materials Forum III, Materials Science Forum, Trans Tech Publications Inc., ISBN 0-87849-402-2, Vol. 516, pp. 1163-1167. RSA 1983, Regulamento de Segurança e Acções para Estruturas de Edifícios e Pontes, Dec. – Lei n.º 235/83. Casa da Moeda. 31 May (in Portuguese). Branco J., Cruz P., Varum H. and Piazza M., Portuguese traditional timber trusses: static and dynamic behaviour, Report E-19/05. DECivil, University of Minho, 2005, 50 pp. EN 12512: 2001, Timber structures - Test methods – Cyclic testing of joints made with mechanical fasteners, CEN, 2001, Brussels, Belgium. Parisi, M.A. and Piazza, M., "Dynamic modelling of friction joints in traditional timber structures", Proceedings, EURODYN 99 Conference, Prague, 1999. Fumanelli, M., Analisi teorico-sperimentale del comportamento sismico di coperture lignee traditionali, Graduation Thesis, Curriculum in Civil Engineering, University of Trento, Italy, 2001 (in Italian).
which requires often an assessment of the actual safety conditions and monitoring. UM. together with the main objectives of the currently running research.uminho. These studies are briefly described and the major results of the finished research studies are presented. case studies of monuments in Portugal including roof and floor structures are briefly presented. In particular. a Church in Coimbra and a Church in Braga. Research studies carried out in UM and currently running on LNEC and UM are also presented. In the present paper. Introduction Masonry and timber are traditional building materials that still face wide application today in modern building industry. PhD Student since February 2005 with the thesis "Safety evaluation of ancient timber structures" in UM.pt Paulo LOURENÇO Assistant Professor University of Minho Guimarães Portugal José SAPORITI MACHADO Forestall Engineer Assistant Researcher LNEC Lisbon Portugal Ricardo Brites is a research Assistant since September 2004. strengthening. the rest of the case studies have been solved with non-invasive and traditional techniques. in which replacement of the structure resulted from the owner decision and extensive damage. Summary The present paper presents some of the recent activity of University of Minho. In general. regarding case studies for Portuguese timber structures in four ancient buildings. which requires structural analysis and. including the Cathedral of Porto. it is also often the case that higher live loads are applied due to a change of use. University of Minho has been involved as a consultant in several case studies in the last five years. NDT is combined with analysis methods aiming at non-invasive strengthening solutions or replacement of the timber structure (only as a last resort). a large number of timber structural systems remain active today in Portugal (even if a significant part of roofs have been replaced in the 20th century by light reinforced concrete slabs). the association of a double roofing system with a masonry vault (for fire protection) and timber roof with clay tiles (for water protection) is common in several European monuments. He has cooperated in several works with Masonry and Historical Constructions Group. Most ancient constructions adopt building systems in timber either for floors or roofs. Portugal November. being showed the relevance of this research studies for the assessment and analysis of historical timber structures. In the case of floors. With the exception of the first case study.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 1. if necessary. The process of assessment and analysis of Portuguese historical timber structures were improved by the studies developed in UM and currently taking place in LNEC and UM. 2006 Selected research and case studies in ancient Portuguese structures Ricardo BRITES Research Assistant University of Minho 4800-056 Guimarães rbrites@civil. together with conservation and remedial measures. 17 .
Coimbra The origin of this monastery dates back to 1131. see Fig. 2006 2. see Fig. Larger defects are also present. (d) detail of vault structure The roof timber structure is extremely complex as a result of additions and successive attempts to correct the observed damage.1 Reliquary room of Santa Cruz. 3. see Fig. with a four-sloping roof with clay tiles and a timber lathwork vault rendered with stucco. due to deformation of the roof above it. The dimensions of the room are 12. see Fig. 2b. Two solutions have been proposed to the owner: (a) keep the existing timber structure using new 18 . see [1] for details. 1 Reliquary Room: (a) plan with main vault and secondary vaults. 1. see Fig. Damage includes humidity stains and rotten roof sheating (the estimate is that 30% of the sheating needs replacing). struts and wedges supported in the timber vault have been added to the structural system. (c) view of lathwork vault. The reliquary room exhibits considerable deformation and cracking of a timber vault. The thinner elements (sheathing) are usually strongly deteriorated. The vault ribs are widely attacked by the common beetle. usually limited to the sapwood. see Fig. 2b. The results indicated that only the sapwood of the structural elements is attacked and the supports of the timber logs are in good condition. d. Reinforced concrete ring beam Timber structure (a) (b) ribs Cambotas Fasquiado lathwork Estuque stucco (c) (d) Fig. The wood elements are generally attacked by xylophagous insects. As a result of the deformation of the truss elements. Selected Case studies 2. (b) main roof structure. 2c. (c) Resistograph to inspect the density profile of wood and the quality of the inner part of the logs [3]. Several truss elements exhibit excessive deformation. together with inefficient connections and corroded ties. the following techniques were used in combination with visual inspection and wood tapping: (a) Pilodyn to characterize the strength to superficial penetration [2]. which is leading to cracking and damage of the lathwork vault.5 x 15 m2 in plan. Portugal November. In order to assess the timber condition. 2a. (b) a chisel to inspect internal surfaces.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.
It is stressed that the presence of the inner vault does not allow a traditional roof structure using a tie beam. plus an ONDULINE type waterproof sub-roof system and clay tiles. the solution includes four diagonal rafters and a ridge board. (b) moisture. together with a set of purlins at mid-height of the rafters. The owner did not want to keep the existing roof structure due to the fact that previous remedial measures did not provide a solution and the vault was in bad conditions. two central trusses and four side trusses. 2006 curved elements supported on the side walls (preferred solution) or (b) adopt a new timber structure. 4. Therefore. aiming at reducing the span of the main truss elements. (c) excessive deformation and rotten wood. (a) (b) (c) (d) Fig. Portugal November. a new roof structure was designed using pine sheating and pine truss elements. The supports of the diagonal rafters and the side trusses 19 . (d) defects Fig. see Fig. 3: Details of the generalized infestation and the irregularity of the cross section of the ribs due the attack Therefore. 2: Aspects of the roof: (a) complexity of the structural system.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.
The airflow will be ensured by small plastic tubes located in the bottom part of the roof.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Portugal November. sheating and strips). 5. (b) sections AB and CD. (c) detail of the central truss support. All steel elements are made using stainless steel type AISI 316. The supports of the central trusses will be made in steel corbels connected to an additional beam. (d) steel beam for central truss support. The top ridge tiles are to be placed dry. The final aspect of the new structure is shown in Fig. 20 . 2006 will be made in existing recesses in the ring beam present in the masonry walls. (a) (b) (c) (d) (e) Fig. (e) 3D model of the new roof The steel connections of the new timber structural system are nailed or bolted with steel plates. 4: Details of the replacing roof structure: (a) structure (plan. so that the bending moment in the masonry walls is reduced. in order to allow ventilation.
and (old) lathwork. above. strengthening and local replacement. consolidation.2 Cathedral of Porto. 5: Final aspect of the new timber structure (a) general view. application of preservation products. Portugal November. With respect to waterproofing and roof structures. (b) new structure. see Fig.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Remedial measures included cleaning. the works aimed at using traditional techniques and keeping remedial measures to the minimum. below 2. 2006 (a) (b) Fig. (a) (b) 21 . application of biocide. The decision was based in continuous in situ diagnosis and structural assessment. 6. Porto The origin of this building dates back to the middle of the 12th century and large conservation works have been carried out recently.
so that contact with the floor structure was impossible during the loading process).International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. the structure would not be able to withstand the requested live load. 8.06x0.024 m 0. (b) structural details and local replacements. Before carrying out the load test. It was possible to verify that the timber is globally in good condition and exhibits no structural damage. see Fig. 7: Arcos de Valdevez City Hall's floor structure: (a) plan. (c) clay tiles details 2. upon doubts about the load bearing capacity of the floor structure. Portugal November.3 City Hall.075 × 0. an in situ load test has been carried out.70 m. Barrotes de m Transverse beams.49 m 0. During general rehabilitation works of the building.10 m Betão leve Light concrete Placa de aglomerado de madeira MDF board MDF Vigamento principal Main beams Vigamento secundário Secondary beams Vigamento principal Main beam system. each 0. Therefore. On top of the transverse beams. each 0. 22 .5 m. The oak timber structure of the floor has a span around 6.075 m2 0. Arcos de Valdevez This case study resulted from works non-carefully carried out in the audience room of this City Hall. placed each 0. (c) a set of transverse beams.70 m m afastado de 0. 2006 (c) Fig. Such additions multiplied the self-weight of the floor around three times (per square meter).46 m. The load test was carried out for a live load of 3 kN/m2. (b) secondary beams located only in the area closer to the main façade.70 (a) (b) Fig.06 m0. (b) cross-section The structure was visually inspected and tapped. being composed by: (a) a set of main beams.46 mde 0 afastados Secondary beam system Vigamento se h 0. Afterwards.20 m b B H Barrotes de madeira Timber beams with com m2 0. a 24 mm thick MDF board has been laid and a 10 cm thick light concrete levelling layer as also been added. placed each 0. a timber board and a levelling light concrete layer have been added to the floor structure without any consideration about possible structural strengthening and composite action. 6: Typical roof details: (a) roof during and after works. the structure was analysed and it could be concluded that. see [4] for more details. a safety system using temporary propping was required (minimum distance between props and floor was 50 mm. even if treatment against xylophagous insects in needed. without the floorboard.075 m 0.20 m 0.
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
The load test was carried out according to [5], given the lack of national or international normative. The load was applied using a flexible water reservoir without bottom, so that no favourable action from the deposit would affect the structure, see Fig. 8. Nine displacement transducers were used to control the test.
0.50 m
0.50 m
Loading Área dearea ensaio
Fig. 8: Load test: (a) water reservoir for load application; (b) detail of fixings for vertical displacement transducers; (c) location of reservoir; and (d) location of displacement transducers The load test was carried out in four loading steps: (a) 50% of the total load; (b) 100% of the total load; (c) unloading up to 50% of the total load; (d) total unloading. Each load step lasted between 25 and 40 minutes, making a total test duration of 5 hours, including the time intervals for loading / unloading decision. Upon completion of the test, residual displacements were observed in all displacement transducers, which is normal for any in situ load test. Fig. 9 illustrates the flexural response of beams V2 and V3. The maximum displacement measured in beam V2 is 18.3 mm, value that should be lower than 1/300 of the span for short term loading (21 mm). The average residual displacement is 9%, which is more than reasonable for this type of structure. Therefore, the floor structure can be used for a load of 3 kN/m2, without strengthening.
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
1ª Fase 21.0 Deformação em mm Displacements (mm) 18.0 15.0 12.0 9.0 6.0 3.0 0.0 11:16
2ª Fase
3ª Fase
4ª Fase
Viga V3 Viga V2
Bea m V2 V3
Maximum displacemen t 18.31 mm 13.93 mm
Residual displacemen t 7.15% 10.56%
Loading area Tempo
Fig. 9: Load test results for beams V2 and V3: (a) diagram of vertical displacements vs. time; (b) maximum displacements at mid-span and residual deformation
Our Lady of Conception, Braga
The present study aims at assessing the safety conditions of a floor structure in a church subjected to a change of use (from a choir not opened to the public in general, to a museum), see [6] for details. The floor is located in the Church of Our Lady of Conception dates from 1625, see Fig. 10, and exhibits considerable permanent deformation (sagging).
(a) Fig. 10: Choir floor's views: (a) from below; (b) from above
The span of this floor is large (around 9.15 m) and the structure is composed by the following elements: (a) a set of main beams, placed each 3.40 m; (b) a set of secondary beams, perpendicularly to the main beams, with a distance of 0.50 m; (c) a set of smaller cross-beams that provide support for the floorboard; (d) a set of steel props, located at one-third span of the main beams, see Fig. 11a. The main beams have a square cross-section with a size of 300 mm, as shown in Fig. 11b, whereas the secondary beams have a rectangular cross-section with 65 × 100 mm2 (see Fig. 11c). The crossbeams are also square, with a cross section of 65 × 65 mm2. The floorboard has a thickness of 30 mm. All structural elements are in chestnut, with the exception of the cross-beams, in eucalyptus and more recently laid. The floorboard is partly in chestnut (original and turned upside down from a previous repair) and partly in pine (more recent). A visual inspection was combined with tapping, in order to assess the condition of the timber. In addition, the resistograph was used close to the supports of the main beams. In general, the
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
structural elements are in good condition, with the exception of one beam, see Fig. 11a. The floorboard close to the walls has a considerable reduction in the effective cross-section due to active biological attack. The biological attack in the structural elements is mostly superficial.
Secondarysecundário beams Vigamento Cross beams Travessas
(Igreja) (Church)
Steel props Escoras metálicas
Main beams principal Vigamento
30 mm 500 mm
500 mm
500 mm
150 mm
65 mm
300 mm
300 mm
(b) (c) Fig. 11: Timber structure: (a) plan; (b) cross-section of the main beams; (c) cross-section of the secondary beams The floor structure was then analyzed according to EC5. The structural analysis indicated that the secondary beams, the main beams and the steel props are not enough to support the code live load of 3 kN/m2. Therefore, strengthening is necessary. The aspects taken into account for strengthening design were: (a) keep the present floor configuration and preserve the existing materials; (b) economy; and (c) do not remove the polychromatic sheating under the floor, due to its artistic value. The following strengthening solutions were suggested to the owner: (a) strengthening of the secondary beams, using a new set of timber beams in between the old beams, and addition of a steel profile for strengthening the main beam as a steel-timber composite beam (the steel props would be removed and the main beams would be forced back to the original position); (b) strengthening of the system by new steel columns (replacing the existing props) and, again, adding a new set of secondary beams.
100 mm
65 mm
100 mm
440 mm
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
600 550 500 450 400 350 300 250 200 150 100 50 0 0 20 40 60 80 100 120 140 160 180 200 220
380 [mm]
(c) Fig. 12: Details about timber condition: (a) near support of a damaged beam; (b) biological attack at the floorboard; (c) Resistograph results
3. Recent research in ancient timber structures taking place in U.M. and L.N.E.C.
In the last years, the Masonry and Historical Constructions Group of University of Minho has made some research studies, aiming a better and profitable assessment of ancient Portuguese timber structures, namely NDT correlations with destructive tests and structural behaviour. It were tested several NDT correlations to chestnut wood (Castanea sativa Mill.), see [7] for details, due to its wide use in Portuguese historical timber structures, namely in the northern region of Portugal. In fact, the chestnut became dominant and indispensable for the survival of mountain populations, and this wood was considered as a noble wood, being used in palaces, castles or in the interior of churches. In the south of Portugal, Pine predominates. The main results obtained in these studies will be now briefly described the and will be presented the research projects actually running in this area in University of Minho (UM) and National Laboratory of Civil Engineer (LNEC). 3.1 Non-destructive and destructive tests in new and old chestnut wood
Feio [7] made an experimental set-up in order to establish the mechanical properties of chestnut wood (Castanea sativa Mill.) by destructive and non-destructive tests in small clear test specimens, for new and old chestnut wood. Chestnut was chosen due to its wide use in historical timber structures in Portugal and the lack of knowledge about the mechanical properties of the old chestnut wood, used in some historical buildings.
with different mechanical properties. the chestnut wood has not suffered any mechanical damage during his lifetime. the safety assessment of historical chestnut structures can be analysed with the same mechanical properties of new chestnut wood. Portugal November. the results indicated that a clear definition for the conventional 27 . This work involved about 370 destructive tests (160 for compression perpendicular to the grain. for small clear test specimens. 13: Test specimens prepared by Feio[7]: (a) for compression perpendicular to the grain. and therefore. 13.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. being compared destructive tests results and a FE model created for this purpose. Resistograph. (b) for compression parallel to the grain and (c)for tension parallel to the grain Carried out tests allows to conclude that density has a significant influence on wood mechanical properties but cannot explain their variability and should not be relied upon as a predictor for mechanical properties. namely Pilodyn. in order to formulate a sensitivity analysis. old chestnut wood (OCW). see Fig. Hence. In addition. and the remaining. There were also made a set of non-destructive tests for correlation with the latter ones..e. and Ultrasonic tests. 2006 The main result obtained of this study is that the mechanical properties of the old chestnut wood (removed from an existing structure) are not worst than the new one. (a) (b) (c) Fig. The load carrying capacity of traditional timber joint was also evaluated in this work. half of these new chestnut wood (NCW). 94 for compression parallel to the grain and 114 for tension parallel to the grain). the mechanical properties obtained by destructive and non-destructive tests in old chestnut wood produces equal or even better results than the same tests made in new chestnut wood. because the correlations obtained between density and mechanical properties are rather poor. i.
the Resistograph and the Pilodyn provide reasonable correlations. (c) Density vs.0 vs. In particular. according to Feio [7] Novel single-parameter linear regressions have been proposed for density. (b) Density vs. Finally. (a) (b) (c) (d) Fig. As this is not reasonable for practical purposes. The usage of the Resistograph and the Pilodyn to obtain quantitive mechanical data is not recommended. 2006 compressive strength perpendicular to the grain is required. 1. 1 – Experimental and Normative values for tension perpendicular to the grain. resistograph measure. Tab. and used as a design basis.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. defining the geometry and loading conditions. with respect to density. 28 . expressions with a lower 95% confidence of the mechanical parameters have been proposed for the ultrasonic testing. elasticity modulus and tensile/compressive strength parallel and perpendicular to the grain. The global conclusions are that. (d)Ec. must be developed and standardized. angle. since difficulties where found when using the Brazilian standard NBr 7190 (1997). using the Resistograph. obtained by the indirect method. Pilodyn depth. the characteristic value proposed by the EN 384 [8]. see Tab. due to the high dispersion found. Pilodyn and ultrasonic testing. an appropriate test method for tension parallel to the grain. Portugal November. Edin. seems to be unsafe and unable to provide a true indication of the compressive strength. correlations need a re-calibration with the wood population. With respect to mechanical characteristics. 14: Some of the key results obtained by Feio[7]: (a) compression perpendicular to the grain vs.
Even if. In addition. (c)adopted constitutive model for FE analysis. The experimental analyses allowed a better understand of the behaviour of the connection and discuss the influence of defects in the mechanical behaviour. 15. 2006 The behaviour of a traditional mortise and tenon joint has been investigated experimentally and numerically under monotonic loading conditions. These models show that the influence of the interface stiffness between the two timber members is significant for the response. (b) photograph of the test apparatus. the usage of non-destructive testing confirmed the good correlations between ultrasonic testing and the ultimate strength of the joint or the stiffness of the joint. such as the normal and tangential stiffness of the interface. and a good agreement has been found between the numerical results and the experimental data. (d) numeric model The investigation led to an improvement of the global structural analyses of traditional timber structures. and was shown to be effective in the prediction of strength and stiffness properties of the tested specimens. Finally. 15: Behaviour of the traditional mortise and tenon connection: (a) experimental set-up developed by Feio [7]. providing valuable information for the transition from classical schemes (hinged or fixed joints) to more sophisticated analysis with semi-rigid joints. the elastic modulus of elasticity and the compressive strength. can be tested using a numerical approach. The proposed compressive constitutive material model (anisotropic Rankine-Hill composite yield criterion. The sensitivity of the results to the various material parameters. see Fig. the failure modes and the better understanding of the behaviour of the joint provide a clear guidance for strengthening strategies in this type of traditional timber joints.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. the usage of the 29 . with different strength values for tension and compression) improved the predictions of local and global behaviour. (a) (b) (c) (d) Fig. Portugal November.
International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. based on that. The decay produced by fungus is very similar to the reduction of mechanical properties due fire in timber structures. adequate condition survey of existing joints using non-destructive testing is possible. The apparatus to create decayed test specimens is to place them into a tube filled with an optimum environment for the development of fungus. a relationship between time of exposure and remaining load carrying capacity. 2006 novel correlations in engineering applications is questionable due to the small sample of joints tested. together with the fungus itself. in order to establish the influence of the area/perimeter ratio in the results. the results indicate clearly that. the main idea is to develop a method based on that one. LECA. There will be used pine wood (Pinus Pinaster ait. but this is not useful for practical situations). and the mechanical properties will be established by destructive and non-destructive tests. 3. Hence. in test specimens subjected to several exposure periods. In order to overcome this issue. The aim of this study is to develop a relationship between the exposure time to fungal attack. it will be calculated the rate of penetration of fungal attack per unit of time and. the depth of that attack and the remaining load carrying capacity. (a) (b) 30 . Portugal November. for correlations. 16 is shown come decay columns actually mounted in LNEC. because there are no direct prediction models for the decrease of mechanical properties due decay (the models actually used correlates the loss of mass with the loss of mechanical properties. This problem is even grater if the structure has suffered from decay. in order to allow a prediction of the load carrying capacity. it has been started in National Laboratory of Civil Engineering (LNEC) a set of experimental studies regarding the definition of a model to help the survey of the remaining strength of existing structures.2 Modelling of the decay of timber structures One of the main problems regarding the structural survey of timber structures is the assessment of the mechanical properties of their elements. therefore. due to it wide use in Portuguese timber structures. The accurate determination of the depth of the attack will be made using FTIR analysis to define the rotted area.) and chestnut wood (Castanea Sativa Mill. In Fig. composed by sawdust. vermiculite and malt extract. It will be used several cross-sections.) in these tests. which have suffered a fungal attack. at least.
2003. L. 1994. today's practice is back to conservation of existing timber structures. Report 03-DEC/E-1. Zombori. In the case of monuments.net – March 2001. Verificação da estabilidade de um pavimento da Câmara Municipal de Arcos de Valdevez. The research that currently takes place in LNEC will give a step further in the assessing of historical timber structures that have some decay problems.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.. In the case of historical centres (vernacular anonymous architecture).. (d) incipient decay detected in some test specimens 4. Proceedings of Pacific Timber Engineering Conference. B. Portugal November. inspecção e reforço do pavimento do coro da Igreja da Conceição. University of Minho. the type of wood (hardwood) and the quality of the construction make conservation possible in most cases. Resistographic inspection of construction timber. Ramos. allowing a more reliable knowledge of his behaviour and hence minimizing the intervention and the respective costs. 6(3). poles and trees.pt/masonry) 31 . Diagnóstico. Spain. Australia. make impossible such a conservation approach. Gold Coast. L.B. Report 03-DEC/E-24.. NDT... 5.B.. (c)Detail of the mixture used to fill the tube. P. 16: Decay tests currently running on LNEC: (a. PhD Thesis. b) mounting of a decay column. "In situ" nondestructive testing of built in wooden members. Rinn. that represents the majority of them. 1998.. Lourenço. After a period in the recent past.O. sometimes the combination of heavy deterioration and original low cost construction. A. Lourenço.F. A. Conclusions Different case studies regarding ancient timber structures in Portugal have been presented. inspection techniques and structural analysis make possible such an approach. University of Minho. 2006 (c) (d) Fig. Inspection and diagnosis of historical timber structures. Ramos. (Available in www. 2006. The recent studies in historical timber structures that is taking place in UM provides a valuable contribute for the assessment of existing Portuguese timber structures.F.J. University of Minho. F. References [1] [2] [3] [4] [5] [6] [7] Feio.. University of Minho. P.. Feio.civil.uminho.B. in which timber structures have been replaced by steel and concrete structural systems. Pruebas de Carga en Puentes de Carretera. Coimbra. Projecto de consolidação da Sala do Relicário do Mosteiro de Santa Cruz. 2003. 2000. Lourenço. Report 03-DEC/E-14. 2003. The combination of knowledge. MOPU. em Braga. P.
prEN 1995-1-2 .International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Office for Official Publications of the European Communities. Brussels. European Committee for Standardization. Belgium. Portugal November. 2003 .Part 1-2: General Structural fire design. Brussels. Final Draft. 2006 [8] [9] CEN. 2004 –EN 384 – Structural timber – Determination of characteristic values of mechanical properties and density. Belgium.Eurocode 5: Design of timber structures . 32 . CEN.
President of GECORPA. is composed of approximately sixty blocks. 1070-296 Lisbon vitorcoias@gestip. particularly after the onset of reinforced concrete. New materials and techniques can be applied in innovative solutions to rehabilitate historic buildings by improving the global strength. restoration of timber and other rehabilitation techniques.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.º dto. SA Rua Marquês de Fronteira. 2006 Rehabilitation of Lisbon's old "seismic resistant" timber framed buildings using innovative techniques. 3. including of architectural heritage projects. FRP. One of the most remarkable features of the buildings which make up this area is their structural concept. Director of "Pedra & Cal" magazine. carried out in a piecemeal process. 8.º dto. walls. also known as Baixa Pombalina. 1. Since 2003. Raquel PAULA MSc in Structural Engineering (IST) R&D Projects Stap. timber. therefore. ductility and energy dissipation capacity of the buildings whilst respecting their original structural concept and. Introduction Lisbon's "Baixa Pombalina". SA Rua Marquês de Fronteira.pt Training period in structural design.pt Vítor CÓIAS Graduated in Civil Engineering (IST) President Grupo Stap. Within each block. is relevant as urban and architectural heritage and therefore worth preserving. But it seems reasonable to admit. Invited lecturer in seminars and posgraduation courses. Founder and president of a group of SMEs dedicated to structural rehabilitation. 8. 3. Keywords: rehabilitation. The gaiola was a standardized construction system which incorporated a set of features designed to provide the buildings with adequate seismic behaviour. beams. enabling them to resist horizontal loads and to dissipate substantial amounts of energy. their authenticity. Structural specialist of the Portuguese Engineers Society. Assistant in FCTUNL and in LNEC. sharing the same gable walls. Some of the blocks have suffered. MSc thesis on confinement of RC columns with CFRP. As often as not this type of buildings requires interventions in respect of the original structure. Summary The historical downtown of Lisbon. gaiola. most of them rectangular and consisting in average of seven buildings. It is desirable (except for safety reasons) that no modifications are now allowed in the blocks which survived until the present day with the original Pombaline structure mostly unchanged. connections. Some research work regarding fibre composites. building by building. deep structural modifications. 1070-296 Lisbon raquelpaula@gestip. that new alterations can be introduced using modern materials and Page 1/13 33 . which was used in the reconstruction of Lisbon after the devastating earthquake of 1755. the distinctive gaiola system. the historical downtown rebuilt after the disastrous 1755 earthquake. Portugal November. in the already modified blocks. is responsible for the R&D + Innovation Projects of Stap. the buildings are constructed side by side.
International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Here "Pombaline" is the term coined after the Marquis of Pombal. the historical downtown of Lisbon rebuilt after the great earthquake of 1755. Further bracing is provided by the Page 2/13 34 . Among these measures. Manuel da Maia. Fig. Characterisation of the "Baixa Pombalina" 2. the existing architecture. Building with the façade walls removed to show the timber bracing system. "Baixa Pombalina". are not at all suited for this purpose [1]. enabling them to resist horizontal loads and to dissipate substantial amounts of energy. 1 Part of a Pombaline block. presenting low intrusiveness and. Methods based on the use of advanced materials and new techniques seem much more suitable. The system consists of a set of timber members embedded along the inner face of the main stone masonry façade walls (Figure 2). the prime minister at the time of the 1755 earthquake.1 The Pombaline buildings In line with the post earthquake measures designed by the chief engineer. Some research work which has been carried out is briefly described. with only tolerable disruption of the buildings. the so-called "gaiola" or "cage" system stands out. the distinctive "gaiola" system. maintaining. The inherent low intrusiveness is underlined. at the same time. resorting to current technologies and materials (cement and reinforced concrete). an internal timber grid is connected by means of iron cross ties. reversibility. The rehabilitation methods used so far. regarding particularly the strengthening of the original "gaiola" framework and its connections to the main masonry walls. To these members and to the ashlars around the openings. who took most of the decisions regarding the reconstruction of Lisbon [1]. introduced after that earthquake (Figure 1). is relevant as urban and architectural heritage and therefore worth preserving. as a result of renovations sought by the owners. Portugal November. therefore. interventions in full respect of the original structural concept. This type of edifice requires. 2006 technologies. the new buildings during the first decades of reconstruction after the big shake incorporated a set of features intended to provide them with adequate seismic behaviour. One of the most remarkable features of the buildings which make up this area is their structural concept. which will enable structural rehabilitation interventions to be carried out "surgically". either in terms of their users or in terms of their structural system. 2. whenever possible.
instead of rubble. being constant for all the buildings. But the buildings of the historical downtown are grouped in rectangular blocks. it seems that such a code never existed in writing. 0. However. It is said that the building concept was implemented by one of the architects. These details were probably handed down to the building carpenters orally or by means of drawings and sketches and were soon lost. named "hands". to a "know-how" that lasted. They gave rise. presently. the buildings needed to have four floors instead of two. establishing member sizes. The confined facade pilasters are then connected to a two-directional vertical bracing system of timber framed walls with light ceramic and rubble masonry infill. namely. bolted to the floor beams and deeply embedded in masonry main walls. the adoption of timber framed walls with mortared stone infill. using units of soldiers to generate the dynamic actions. including ground floor and attic and the height of the façade was approximately equal to the width of the main streets. Italy. 2006 timber floors. Fig. The "gaiola" system is considered an innovation of the Pombaline city planning. when the last "gaioleiros" (buildings using the system) disappeared [1]. The main walls are connected transversally at ground level by other masonry walls. it is assumed that most of the constructive details of the new structural system were produced by the master draughtsmen of Casa do Risco (Draughtsmen House).9 m thick at the ground level. regulating design. But. the concept seems to have been inspired on the Pombaline constructional system [1]. Structural elements of floor slabs. Portugal November. These walls are around 0. whose diaphragm action is enhanced by iron ties. Notwithstanding the many references to a possible code. Floor coverings. Wooden floor slabs. dated from 1958.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. connection details and constructive procedures. In fact. adopted when it was found out that. 2 Details of the Pombaline construction. and by timber connectors. with a slight reduction in the upper floors. used to the design of timber structures for ships. In the city council archives. Records on the original design of the Pombaline buildings seem not to be available. as was Manuel da Maia's initial idea and was the case at Vila Real de Santo António. starts with a reference to "the code published after the earthquake of 1755" but. The "gaiola" designation was coined because the building seemed like a big cage. after a load test on a large scale model in the main downtown square. appears often in specialized literature as having given rise to the first rules for seismic construction. meanwhile. Load bearing composite wood-masonry walls. with the carpentry work high up in the air [1]. nailed to the above mentioned timber grid and also embedded in the masonry. only documents pertaining to the successive modifications requested by the owners are available. in spite of suffering substantial degradation. in the Algarve. The buildings had originally five floors. unfortunately. no document has been found specifying the "gaiola" construction or establishing its mandatory character.5 m thick. until the first quarter of the 20th century. with a narrow central yard measuring 45 u 2 m2. measuring in plan around 70 u 25 m2. the earthquake of 1783. Page 3/13 35 . in Calabria. Internal timber frame. due to complicated property expropriation and reassignment. The first modern Portuguese seismic code. Each block has stone masonry walls in the external walls and in the internal walls facing the central yard. Carlos Mardel.
International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. both in the façade and in the back yard are strengthened as detailed above [1]. The structure of the first floor is generally made up of stone masonry arches and vaults. An example of a representative wooden roof structure is illustrated in Figure 4. b) king post truss. Fig. Portugal November. The empty spaces of the walls were filled with rubble masonry made of small stones and ceramic elements recovered from the ruins of the earthquake. assembled with lime mortar [2]. Timber species used in these buildings and their cross-section dimensions also presented some variation. the pilasters of the exterior walls. Jointed timber elements form the triangular frame. masonry blocks). The foundations rest generally on a timber grid laid on short timber piles. footings. separating the different ownership and acting as fire walls. Page 4/13 36 . Several basic configurations of timber-masonry load-bearing walls have been identified in the old Lisbon buildings (Figure 3). From the first floor up. The rigid structure that supports the roof consists of a set of king post trusses. intended only to stiffen the alluvial soil and to create a good working platform at the ground water level. 2006 Some of these walls are continuously extended further up above the roof. Roof structures also include the secondary elements underneath the roof coverings. In many places where the new buildings are laid out on the remains of pre-earthquake buildings (street slabs. Fig. 4 Details of the Pombaline construction: (a) roof structure. The king post is the central vertical timber. 3 Examples of different configurations of timber-masonry walls. these older constructions were used as foundations.
these improvement solutions are capable of enhancing the global strength. The introduction of new materials. a) b) Fig. 2006 2. Together. resulted in uncontrolled changes in the original structural system. While in some cases. 3. insect damage. and addition of steel and reinforced concrete elements [1]. Portugal November. possibly decreasing its strength and capacity to dissipate the energy associated with seismic actions [2]. Overloading and modifications on the buildings. by improving the structural behaviour of: the original timber floors. Maintenance and conservation of built heritage frequently involves rehabilitation and consolidation of certain components or parts of the existing timber elements. and the connections between the timber framed walls and the main masonry walls. there is room for low-invasive strengthening solutions. 5 Deterioration of timber floor beams in Lisbon's downtown: (a) decayed beam end. due to their strong invasive character. Current repair methods are often intrusive and lead to the waste of good material. given its the historical and architectural value. to preserve the original materials and Page 5/13 37 . ductility and energy dissipation capacity of the buildings whilst respecting their original structural concept and. Innovative solutions to improve the structural behaviour New materials and techniques can play an important role in the rehabilitation of the "Pombaline" buildings compound. involving tolerable disturbance for the users of the buildings. notably on the ground floor. inadequate widening of existing façade openings and removal of walls and pillars. and biological damage (rot). environmental effects.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. their authenticity. the original timber braced walls. (b) reduction of cross section in middle span. in other situations. Therefore. with mechanical behaviour considerably different from the original ones. mould. of "surgical" nature. In particular. can severely compromise the integrity and performance of wood structures. resorting to current technologies based on the intensive use of cement and reinforced concrete. the connections between the timber floors and masonry walls. the substitution of the original timber is inevitable. Main changes have been the construction of extra floors. Many structures are in need of some kind of rehabilitation due to the deterioration they have undergone with time. the interventions need to have minimum environmental impact both on the existing use of the building and on the appearance of the structure. significant reduction of section and excess of deformation are the most frequent anomalies that have been detected on the timber elements (Figure 5). therefore. Decayed zones. that will simplify interventions. on the one hand.2 Vulnerability and deterioration of timber structural elements Buildings in "Baixa Pombalina" have undergone successive and occasional modifications until quite recently. The rehabilitation methods used so far. on the other hand. are not at all suitable in the case of the "Baixa Pombalina".
International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Portugal November. versatility and structural efficiency.1 FRP reinforcement FRP composites combining high resistance fibres (glass. epoxy adhesives are in general terms the most suitable for in-situ repair operations. as well as concrete and masonry walls that lacked adequate steel reinforcement. therefore suitable formulations should be identified for each job. strength and elasticity. free-form and tailored design characteristics.2 Low intrusion restoration techniques for timber structures Low intrusion restoration techniques using structural adhesives and FRP are suitable for repairing and strengthening existing timber structures and represent significant improvements in current practice in conservation strategies for timber. Epoxy adhesives include a wide variety of products with quite different properties in terms of adherence to timber. FRP techniques should only be applied by certified applicators using certified equipment. End connection details can also be incorporated to assist in load transfer between elements [5]. Key advantages of FRP are high strength/weight and stiffness/weight ratios. New materials can be applied in innovative solutions to improve the seismic behaviour of old buildings. which significantly exceed those of conventional civil engineering materials. flexible sheets or fabrics. In combination with traditional carpentry methods. sporting goods industry. it is better to adopt a local renovation of the decayed parts as well as the reinforcement of the existing structural elements. and can be implemented without the need for foundation reinforcement because the materials are so lightweight. 3. either being inserted in critical cross-sections. ease and speed of installation with minimum personnel and plant requirements. Most of the new solutions can be optimised by the use of nontraditional materials like epoxy resins and advanced Fibre Reinforced Polymer composites (FRP). viscosity. made of fibres in one or at least two different directions. Despite some less good qualities. respectively. 3. The growing use of FRP in repair and retrofit of concrete and masonry structures has opened the door for similar applications in wood. creep. aramid or carbon fibres) and a polymeric matrix (epoxy. Composite materials for civil engineering are available mainly in the form of: thin unidirectional strips and plates (laminates). 2006 structure or even for economic reasons. Unlike traditional methods. reduced Page 6/13 38 . has given rise to their application in the civil engineering sector and are currently viewed as highly promising materials in the construction industry. polyester or vinylester resin) have a wide variety of industrial applications. the new approach is reversible. Innovative solutions are capable of providing the necessary strength and durability to realize a longterm service life for wood structures. corrosion and fatigue resistance and a high degree of inertness to chemical and environmental factors [3]. FRP rods and plates have a particularly effective application for the local reinforcement of structural timber joints. versatile timber restoration solutions lead to less waste of original good material. ease of handling and versatility. FRP composites can be embedded in existing wooden beams and joints to provide a "hidden" strengthening solution. The versatility and ease of installation make FRP retrofit solutions extremely effective [4]. profiles and rods. reaction and setting time. etc. The system can be designed to increase the strength of beams deficient in flexure and/or shear. since they do not require high pressure during their application and curing and they can be reasonably tolerant to glue line thickness variation [1]. It is a less disruptive process. or used for load transmission when damaged beam ends are cut off and replaced with new timber or epoxy mortar. The option for advanced materials aims at providing systems with such features as low mass. The widespread use of FRP composites in the aerospace and defence related sectors. FRP materials have been used successfully to strengthen existing concrete columns for higher seismic loads. These materials are also commonly used in structural repair operations.
7].e. upgrading of glulam beams. They may be one of or a combination of the following: Page 7/13 39 . age. Fig. and prefabricated timber components. i.Low intrusion conservation systems for timber structures [8] . including above decorative ceilings (Figures 7 and 8). Portugal November. beam reinforcement. beam end repair (Figures 9 and 10). based on the experimental trials. if necessary. forming an integral part of the repaired structure. 2006 visual impact and an aesthetically pleasing appearance [6. 7 Beam upgrades using: a) bonded-in rods and b) bonded-in laminate. king post truss consolidation and bonded-in joints. Licons . 6 Modified flitch and upgrade of beams a) a) b) Fig.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. thus achieving better construction and large overall project cost savings.European project addressed the development and validation of these innovative techniques as well as the establishment of a quality assurance methodology for executing each type of repair. Main configurations include: modified flitch and upgrade of beams (Figure 6). strengthening rods or plates. moisture and visual grading. current condition. The restoration components are introduced to the site.. The systems are intended to reduce disturbance and eliminate the waste that is often associated with fully in situ methods. The primary restoration material is the original timber component that is identified during the inspection with regard to species. The basic components of the technology are the adhesives.
mechanical properties of the adhesives should be compatible with those of timber. for example. b) externally bonded laminates. The adhesive should have low surface tension in order to obtain a good spreading of the material. The prefabricated timber components should be of the same species of the timber to be repaired. The reinforcement can either be in the form of bars or a vertical laminate. If a beam has some rot in the centre. The FRP rods or laminates may have some diverse compositions. Adhesives are usually a two or three part epoxy system which is used as the interface between the original timber and the rods or plates. Page 8/13 40 . After curing. or compatible.prefabricated timber component of the same species and moisture as the original timber. The repair technology was chosen for both on-site accessibility requirements (impossible to work from top) and cost analysis. Resin is injected into the holes and the decayed part is replaced by a timber prefabricated component (Figures 9 and 10).single or multiple configuration of metallic or FRP rods or plates as overall strengthening components or as shear connectors. Case study presented in Figure 10 consisted of pine floor beams. However. Beam ends can be repaired by drilling holes in the sound material after the decayed end has been removed. The intervention was completely carried out from below the beams [12]. to be used in crack injection. Individual rods are easier to handle and will adapt to the line of the beam. unidirectional glass fibres on a polyurethane thermoplastic matrix. unidirectional glass or carbon fibres agglutinated on a matrix of epoxy resin. fixing of anchorages in timber. a) b) Fig. the top surface of the beams was not accessible. The rods and plates can be metallic or of FRP materials. if the durability of the original timber is rather insufficient regarding the particular hazard class. in terms of its mechanical properties. which makes it inadequate. 2006 . The adhesives should be specifically formulated for timber engineering. or it is required to increase the bending capacity. always in accordance with the design. but a higher section modulus can be achieved with a laminate [10]. with significant end decay. timber with adequate natural durability or with a selected preservative treatment may be used [11]. Timber should have a moisture content not exceeding 20%. . filling of drilled holes and slots to fix metallic or FRP rods or plates. durability and colour. then an epoxy may be used to bond steel or FRP rods into a slot cut along the centre of the beam from above (Figure 6). 8 Beam upgrade above decorative ceiling using: a) bonded-in rods. The metallic reinforcement elements can be of stainless or steel adequately protected against corrosion. The metallic rods should be threaded or ribbed bars. etc [9].International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Portugal November. thus needed to be repaired. Due to the presence of the existing wooden floor. usually to replace a portion of a component.
3. facilitating significant cost savings and with little disruption to the space below. on the constraints of accessibility and available space to place the prefabricated components. Improving the connections between the timber floors and masonry walls not only increases the resistance of the structure of the floors but also the global behaviour of the structure of the whole building. The option for the best configuration depends. when existing they are usually deteriorated. Portugal November. a) b) Fig. The bars are inserted into Page 9/13 41 . Low intrusive and removable connectors are shown in Figures 11 and 12. Encase of rods in lateral slots in sound timber. particularly in what concerns the seismic action. 10 Beam end repair using prefabricated timber components: a) injection of epoxy resin into the holes in the sound timber. thus no longer suitable to assure the connection. 2006 Instead of being installed in situ in the sound timber. Fig. 9 Beam end repair using a prefabricated timber component with pre bonded-in rods.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. the reinforcement elements can be pre-inserted in the prefabricated components at the workshop. b) injection of top slot. for example. Injection of resin. because of the contribution of the timber floors to the structural stability [13]. Licons techniques permit the rehabilitation and consolidation of timber structural elements without extra weight and total removal of the sound timber.3 Improving the connections between the timber floors and masonry The devices that were used to connect timber floors structure to the resistant masonry walls.
There are different types of connection devices either to be installed in the longitudinal or in the transversal direction of the beams. The set of pieces can be hidden in a small concavity on the wall and then rendered. sock type anchors may be used (Figure 13). 12 Connection devices between timber floors and masonry. the devices have a special anchoring semi-spherical rótula (ball joint) that enables the fitting of important misaligning holes.4 Strengthening of the original timber braced walls The contribution of the structural capacity of the timber framed walls with light ceramic and rubble masonry infill. like the condition of the materials and the connections between the different elements [13]. a layer of glass fabric and epoxy resin composite may be put between the beam and the plate. the plates are bonded to the pavement. Page 10/13 42 . to the global lateral stability of the buildings is easily understood by current structural engineering. 2006 the thickness of the walls and are anchored on the wall and on the timber beams by special devices. 11 Connection devices between timber floors and masonry. the L-shape plates are connected to the beams usually with threaded bolts. Portugal November.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Of course the structural capacity is dependent on diverse factors. In the longitudinal direction. In order to assure the adhesion. 3. Fig. In order to reduce the visual impact of the anchoring. In the transversal direction. As the execution of aligned holes in the thickness of the walls is difficult to assure. Instead of inserting the bars through the whole thickness of the masonry walls. Fig. the semi-spherical rótula and the nut are lodged in a semi-spherical cup with a lid. Longitudinal direction.
externally bonded GFRP fabric may be used as presented in Figure 13. Also. instead of inserting the rods through the whole thickness of the masonry walls. glass fibre fabric stripes were therefore folded around the edge of the panel. This was done by opening one slot in the broken timber member and placing in it one 10mm diameter glass fibre reinforced polymer (GFRP) rod glued with epoxy adhesive. The anchoring devices may be the semi-spherical ones that have been previously shown in Figure 11. Also in this case.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Full scale models of an individual unit of a typical timber frame wall panel were produced and tested in diagonal compression close to complete failure. after the above-mentioned slot had been filled with epoxy adhesive and this left to cure. in the case of extreme joints. which are no longer suitable to assure the connection. In addition to the reinforcing rods. Portugal November. Every test panel was strengthened by repairing both outer timber members of the frame on the shortest sides of the panel. The L-shape plates are connected to timber with threaded bolts. One of the main deteriorations of these connections is due to the corrosion of the ties. Similarly to the previously presented devices that improve the connections between the timber beams and the masonry. After testing. strengthened and subjected to diagonal compression again. The new timber parts should be connected to the remaining sound timber by means of metallic plates and screws or using the low invasive methods already described. the reinforcing bars are inserted into the thickness of the main masonry walls and are anchored on the timber framed wall by special devices. Page 11/13 43 . 2006 After a very careful inspection. in order to avoid non-symmetrical strengthening effect. This was done on both faces. In order to assess the effects of FRP solutions to strength timber load-bearing walls. Each GFRP rod was positioned the nearest possible to the centre of the corresponding member cross section. followed by its substitution with new timber. it may be concluded that it is necessary to remove some parts of the timber frame. glass fibre fabric stripes were glued to the joints where tension conducted to the separation of timber elements during previous testing. several tests have been carried out [14]. Preferably. 3. The fundamental interventions in this kind of walls are thus related to the reestablishment and even the reinforcement of the connection conditions. the distorted test panels were brought back to their initial configuration. Two rods were therefore used for each panel. The glass stripes were as wide as the timber member they were glued to.5 Improving the connections between the timber framed walls and the main masonry walls The internal timber grid of the original structure of the buildings in "Baixa Pombalina" was connected to the main stone masonry façade walls and to the ashlars around the openings by means of iron cross ties. and 25cm long. The strengthened wall panels have presented a very reasonable recovery of strength (between 73% and 127% of the strength of the initial panel) and a very good improvement in their ductility (between 158% and 316% of the maximum diagonal deformation withstood in the first test). a GFRP fabric may be put between the timber and the plate. This can be achieved by means of the installation of a set of low intrusive and removable connectors as shown in Figure 13. sock type anchors may be used as shown in Figure 13.
[4] M. Technical report. Vítor. 2003. ductility and energy dissipation capacity of the buildings. M..7 References [1] Cóias e Silva.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 13 Timber braced wall strengthened with externally bonded GFRP fabric and connection devices to the main masonry walls.. pp. [7] Paula R. "Sistema pouco intrusivo de reabilitação de madeira". International Conference 250th Anniversary of the 1755 Lisbon Earthquake. [6] Cruz H. therefore. 3. Cruz H. Aramid. FRP Laminates and Epoxy Help Support New Loads in an Existing Wooden Gymnasium".569-578. 2006 Fig. which was used in the reconstruction of Lisbon after the devastating earthquake of 1755. Custódio J. Innovative techniques can play an important role in the rehabilitation of the buildings in the Pombaline downtown.. Éditions Elsevier. pp. "Seismic analysis of a heritage building compound in the old town of Lisbon". a brief description of non-traditional materials and rehabilitation methods. Portugal November. February 2004. 2002. European Timber Buildings as an expression of Technological and Technical Cultures. Execução e Controlo de Qualidade da Reparação de Estruturas de Madeira com Colas Epoxídicas e FRPs. their authenticity. Revista Page 12/13 44 . Lisbon. Palmer. [3] Externally bonded FRP reinforcement for RC structures. also known as "Baixa Pombalina". 1 to 4 November 2005. "Using advanced composites to retrofit Lisbon's old "seismic resistant" timber framed buildings". CEB-FIP. and N. the gaiola system. 3. fib Bulletin 14. Ehsani. 19-20. LLC.. "Strengthening of Old Wood with New Technology. Maintenance and conservation of built heritage frequently involves rehabilitation and consolidation of certain components or parts of the existing timber elements. [5] Tyfo Systems for Wood Using Glass.. Larsen. Whilst respecting the original structure concept and. Cóias V. Fyfe Co. 362-368.. The gaiola was a standardized construction system which incorporated a number of innovative methods designed to resist seismic forces. which are capable of providing the necessary strength and durability to realize a long-term service life for timber structures is presented. CIMAD 2004. pp. comprises the most complete collection of the seismic resistant building technology. STRUCTURE magazine. In this paper. these techniques are capable of improving the global strength. [2] Ramos L. and Carbon Fiber Reinforced Tyfo® Materials combined with our Tyfo® Epoxy Matrices. Lourenço P. The system consists of a set of timber members embedded along the inner face of the main stone masonry façade walls.6 Conclusions The historical downtown of Lisbon.. July 2001.
Departamento de Engenharia Civil. João. Edições Orion. [12] Paula R. Faculdade de Engenharia da Universidade do Porto. 1. [13] Appleton. Rotafix Ltd. Setembro de 2003.-Mar. 2006. [11] EN 1995-1-1:2004: Design of timber structures. Portugal November. 2001... Cruz H. [10] Ross P. Jan.Common rules and rules for buildings. [8] www.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Reabilitação de Edifícios Antigos.. et al.º Encontro sobre Patologia e Reabilitação de Edifícios. LNEC. Thomas Telford Ltd. Um caso de estudo".org [9] Resiwood system timber engineering product data. "The use of FRP in the strengthening of timber-reinforced masonry loadbearing walls".licons. Page 13/13 45 . Appraisal and repair of timber structures..º 29. "Reabilitação pouco intrusiva de vigas de madeira. PATORREB 2006 – 2. [14] Cruz H. Part 1-1: General .. 20-21 Março 2006. London. 2002. Ano VII. 2006 Pedra & Cal.ªEdição. N.
it will later be used for multimedia shows. A SHIP FOR THE FUTURE The sea and the future are the thema of Lisbon EXPO' 98. The contractual design code was Eurocode 5. and construction. the contract turnaround time for this exceptional construction made an imperative condition for design. Bocquet. because it is convenient for curved shapes. 5000 cubic meter of sawn timber were employed. P. Additionnal testing and scientific expertise were nevertheless demanded to assess the design of joints. D. J.T. and erection pro cess of this new basis of design for joints had to be performed in real time. able to resist fire for a long time. The choice of EC5 as design code was another condition. and worldwide recognition. Biger4 1 Ets Weisrock. An unusual amount of cutting and drilling works were made by a new computer driven automat with outstanding accuracy and efficiency. conceived by Architect Regino Cruz in association with consultant Skidmore Owing and Merrill (Fig 1 a).P. and the instructions for the glulam cutting and drilling automat have been carried out by the consulting engineer. The 3D model calculations. 4 Bureau Veritas France. Consistent overcosts resulted from the need encountered by surprise to bring important corrections to the contractual design code. fabrication. unaffected by the humidity prevailing on the banks of river Tejo. lightweighted. and naturally beautiful. The basis of selection defined by Atlântico. without delaying the construction. the drawings. a rare distinction.S. J. The constructor of the hall was to be selected through a european call for offers. and a great capacity in design. Paris Summary The 22nd May 1998 is the opening date of the EXPO'98 in Lisbon. the glulam structure and the roofing were finished at time for the preparation of the first show of the Pavillion of Utopia. visitors will remark the Pavilion of Utopia. this glulam hall is 180m in length. the detailing. the society entrusted with the organisation of the EXPO' 98. The erection on site was conducted on chedule by masters of the art. Saulcy sur Meurthe . Racher3. 2 AGIBAT-MTI. 47 . fabrication. This multimedia showhall with a capacity of 12000 to 16000 people (Fig 1 b). Grade C30 spruce from Sweden was used for the glulam. this project represents a valuable enhancement for the image of glulam.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 1. Ecully 3 C. Glulam was chosen for the structure of the pavillion. and the opening of the EXPO'98. and the extra difficulties to overcome in the site reinforcement of joints. Their implementation in the design. Among the other constructions build for this event on the banks of the river Tejo. Reinforcements and complementary resistance verifications were tested out. F. an experience of precision machining. comprised notably a great fabrication capacity. With a modern touch. Quost². Accomodating 12000 people. Portugal November. a great ability in transportation and site erection.Perrin1. Clermont Ferrand. Full size joint testing showed a real need for improvements. Tightly bound to the opening date. Finally. despite the high complexity of the works.U. evokes through its shape the hull of these Portuguese caravels sailing away to discover the New World. One of its main exhibitions is hosted by the Pavillion of Utopia. 2006 THE PAVILLION OF UTOPIA J. and 125m in width. analyzed and finalized on a scientific basis.P.
Figure 1 a: The architect design for the hull Figure 1 b: The arena and the steps 2. 48 . was also accepted. on the 22nd of may 1998. This shape surrounded more closely the inner volume of the steps. No change in the architecture and design would be admitted.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Portugal November. with zinc coating instead of aluminium. AN ALTERNATIVE DESIGN However an alternative design shape proposed by Ets WEISROCK for the shape of the glulam arches was accepted. not extendable and as imperative as the opening date of the EXPO'. Another alternative solution for the roofing. resulting in a substantial reduction of the bending moments in the arches. and a benefit in cost. 2006 The projected schedule for the construction was as tight as possible.
THE CONCEPTION OF THE STRUCTURE The Pavillion of Utopia covers an area of 15000m2. with a cap extending out for 20m on the west side. 2006 Figure 2 : an alternative design for the arches The contract for the construction of the glulam structure and for the roofing was signed on the 20 of May 1996. fabrication. 49 . purlins. The projected planning included 12000 hours for design. The achievement of such a challenge required a special task organisation and schedule. 3. It is egg shaped.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. and 10 months for erection. To carry out and complete in priority the general design with the definition of geometry and the dimensioning would allow to split the process of detailing. and wind bracings set in between three arches at each end of the building (Fig 3 and 4). The structure comprises 16 transversal arches laid with a regular spacing of 9m. two longitudinal arches framing the sheds of the central part. transportation an erection in many subsequences. two years exactly before the opening of the EXPO'98. struts. Portugal November.
The composition of the diagonals is similar. 2006 Figure 3: layout of the structure The arches range from 52 to 115m in span. The upper members have the same composition. which are glued together. The lower members are 405mm thick. except for the top parts which are thinner with a thickness of 220mm thick in two parts. and from 20 to 34m in height.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. they are composed of three parts each 135mm thick. They are lattice arches with two articulations on the supports. 50 . the distance between those members is variable. The lower members are much bigger in depth than the upper members. Portugal November.
Portugal November. to ask the carpenter for a validation by a scientific expert of the design method for the big sized joints . It seems therefore appropriated in our opinion. the current EUROCODE 5 is a standard with an experimental character. The connections between opposite end plates are made with 33mm bolts These connection bolts are concealed in 70mm circular holes drilled on each outer member part. would not stand to be subjected to the uncertainties which are encountered in experimental approaches of design. Obviously such a project bound to a very tight schedule.» Following the advice given by his scientific consultant. Two arches were erected at the moment.T laboratory. This glulam lattice is linked to the supports through to a lower part made of triangulated metal tubes. they include two layers of glulam lattice. the technical controller Bureau Veritas expressed a different feeling about the exclusion of tests in the contract: « Our point of view is more balanced. THE QUESTION OF JOINT DESIGN As the choice of EC5 as design code was beyond any question. joints are made with one or several pairs of inner metal end plates connected to the glulam with 20mm diameter dowels. to have full size joint tension tests made on a diagonal by the C. The suggestion of the technical controller to submit the case to a scientific international committee was accepted by Atlântico at the very beginning of 1997.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. The opposite plates lay in grooves alternately cut on the sides of the central member part and on the corresponding side of the outer part. a few more were under fabrication. 51 .S. The lower members of the arches are bound together with glulam struts. the carpenter decided without delay. 4. any eventual appeal ta full-scale testing was excluded in the contract. The glulam purlins are laid in between the upper members with a spacing of 3. The windbracings are set at each end of the arches layout.U. ln his first report to Atlântico dated on the beginning of october. and steel ties. The portuguese regulations require a five years garantee. 2006 Figure 4 : The carpenter's model Accordingly to the size of the glulam element and to the forces to transmit.4m. and we must take this situation into account. The insurance of the constructor submitted the delivery of this guarantee to the review of the design documents and of the construction by a technical controller . The tests results were available towards the end of December 1996. one at the upper members level and the other at the lower members leve!. they raised a serious question with the Chapter 6-Joints of EC5.
This possibility for joint reinforcements in the existing arches was given by an over design margin resulting from shortened calculation procedures without optimisation. A postprocessor dedicated to the verification of sections according to EC5 has been especially written for this application. wind loads. DESIGN Design loads comprised.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Solutions to the design problem were found and validated in a short time. The design. Three successive series of tests (Fig 5) were carried out very quickly by the C. The calculation model established by the consulting engineer AGIBAT-MTI includes 3 530 nodes and 5980 bars (Fig 6). THE CHANGES lN THE BASIS OF JOINT DESIGN The challenging task given by Atlântico to the scientific committee in January 1997 was to check the effectiveness of the reinforcements proposed by the carpenter. 2006 5. seismic loads (0.T. fire resistance (1 h). on a H. working station. with 62 load cases and combinations.28g with q=1). Fortunately the implementation of the reinforcements on the existing arches was possible and could be delayed without disruption in the construction schedule.P.22 and 0. The total duration of a computer run was not less than 14 hours.S.U. laboratory under the supervision of the scientific committee. final design specifications were issued in March 1997. Portugal November. dead loads. and to define complementary joint design criteria for the project under construction. The tests analyses have been performed in real time. Figure 5 : testing a diagonal end joints 6. fabrication and construction process had continued during this period. exploitation loads. The computations have been carried out with SYSTUS (a calculation code developped by Framasoft). 52 .
450 tons of steel have been employed for the fabrication of about 8000 different pieces. and over 15 000 more have been executed for the purlins. Over 250 000 holes have been bored in the steel elements. Portugal November. Over 70 000 drillings have been performed for the totality of the 16 arches. 53 . purlins. and for ail the other pieces coming in between. 7. these routines have been rewritten during the design process accordingly to the decisions resulting from the full scale tests.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. The drilling has been made with a precision of 1/10mm for more than 3000 dowels per arch. The west cap and the east gable have been made in a classical manner. 2006 Pvt"'J'HS Figure 6: view of the calculation model Excel routines have been used for the design of joints.0 have been used for the drawings. have been cut and drilled with the automat. representing nearly 4000 different pieces. FABRICATION 3 400m3 of glulam have been fabricated. and 9 000 drawing hours for 930 plans. DXF files ready for use by the fabrication automat have been included in the design. struts and bracings. struts. The design process has taken about 1000 engineer hours. Autocad 12. This fantastic work would not have been possible without the automat (Fig 7). All the glulam pieces for the arches. The changes in the basis of design resulted in significant overtime and overcost for the designer. bracing.
The site reinforcement of the arches which were erected before the implementation of the new design in the fabrication needed time consuming and cost effective detail work from a nacelle. Portugal November. Normal transports took three days. The erection was not possible with usual self moving cranes. with a supervision by CTBA. 2006 Figure 7 : the tool of the automat The quality of the fabrication of glulam has been continuously controlled by the carpenter . Figure 8b Figure 8a 54 . The different phases and details of the lifting of the structure are shown Figures 8a to 8h.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Three shifts per day of ten Portuguese and French skilled companions have been busy with the lifting of the structure during ten months. 8. these operations were conducted by a master in this difficult art. not far from Strasbourg to the site of construction in Lisbon has needed over 90 normal trucks and 36 exceptional trucks. the transport of the longest pieces took 15 days. TRANSPORTATION AND ERECTION The transportation from the workshop in Saulcy sur Meurthe. two tower cranes situated in the arena at the foot of the steps were necessary.
both in plane and in height. In Finland there are no major earthquakes and for this reason expertise in this field is not widespread. 55 . EC8 gives the methods to calculate this shear force. however. Experience from North America and Japan shows that wooden buildings can resist catastrophic earthquakes while sustaining only minimal damage. The ductility of the joints is very critical as it also affects the level of shear force to which the wall is subjected.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Many modern timber buildings have even survived showing no visible signs of damage. for example. are new design codes and these may be applied. In the European region. 1. and in such cases. 2006 Design guidance on Eurocode 8 for practicing engineers for timber structures Tomi TORATTI Doctor in Techn. Wooden buildings are usually regular. ductile joints and in general very regular building geometry. and Eurocode 8. Senior Researcher VTT Technical Research Centre of Finland PL 1000 FIN-02044 VTT. design of timber structures. The body forces created by the ground acceleration on the building are converted to a base shear force imposed in both principal directions. For the exportation industry. a simplified modal response spectrum analysis may be used. The structures resisting these lateral forces such as shear walls. An effective way to design for lateral loads. floor diaphragms and anchorages are then designed against this base shear force. Eurocode 5.toratti@vtt. The sources of information for this report are given in the list of references. Portugal November. This report explains the use of Eurocodes in the seismic design of wooden residential buildings. including seismic loads. in the exportation of wooden buildings and building expertise to seismic areas. These shear walls have a high lateral force-resisting capacity and the joints are in general very ductile. The advantage of wooden buildings is based on low self-weight. Introduction Eurocode 8 (EC8) is a modern design standard for the determination of seismic loads and structural details.fi Summary Wooden buildings have a good reputation when subjected to seismic events. Finland tomi. The high performance of plywood shear walls is based on the ductility and energy dissipative characteristics of nailed or screwed joints on plywood in shear walls. in residential wooden houses. modern design codes perform well for earthquakes. The main sources have been Eurocodes 5 and 8 and the STEP lectures B13 and C17. is the use of plywood panels in shear walls. Based on previous experience. design provisions for earthquake resistance of structures. knowledge of EC8 is often important. The Eurocode 5 and 8 versions referred to in this report are the versions given in the references.
For wooden houses vulnerable parts are: the anchorage of the house. in the side of the shear wall where the compression of the stud is applied to the bottom plate. 2. for example. Only limited damages may be accepted with the building staying intact. ductile structures or structures containing ductile joints perform well during seismic events. In addition. which are important in the community (hospitals. This is based on the low weight of timber structures. inertial forces in the structures. Timber houses have a good reputation for performance in seismic events. 2006 The soil movements induced by earthquakes produce vibrations in buildings and. especially compression perpendicular to the grain. the calculation method of seismic loads and relevant load combinations are given Eurocode 8 part 1-2. Timber behaves in a ductile manner when loaded under compression. 56 . However. allow a significant reduction of seismic loads for ductile structures. As in any kind of building it is usually the inadequate structural design or inadequate supervision during the building process that causes the damages induced by seismic events.). To bear seismic loads. General rules . Also. clear layout of timber houses and good lateral stability of the house as a whole. especially when the tension is perpendicular to the grain. (see Buchanan & Dean. The joints of timber structures are normally more ductile than the timber parts themselves and this is. garages). fire station. Timber is brittle in tension. remain functional. General rules for buildings. This reduction takes into account the ability of the structure to deform during seismic events. since in this case only a small deformation may cause failure. 1988 and Ceccotti.2 The structure of Eurocode 8 EC8 is divided into three parts as follows: Eurocode 8 part 1-1. This part outlines the general rules regarding seismic resistance. These possess an ability to withstand deformations without developing high stress concentrations. etc. the diaphragm action of floors and the first soft storey which sometimes has been left without sufficient lateral bracing (for example crawl spaces. The present paper is a short version of a more complete report ref. 2.Seismic actions and general requirements for structures. Buildings built in seismic areas should be designed and built so that there is no danger of collapse. This is advantageous in seismic design as. limited damages and that those buildings. the reason for the overall ductile behaviour of timber houses and their good seismic performance. Most seismic design standards.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. including EC8. the general requirements and definitions of seismic-resistant buildings are stated. such reduction of seismic loads is not allowed.1 Seismic design of timber houses according to Eurocode 8 Introduction Eurocode 8 (EC8) gives instructions on how the seismic loads are to be calculated. Portugal November. perpendicular tension stresses should be avoided. ductility of joints. 2000). This paper deals with only those parts of EC8 which concern timber structures. For brittle structures. Stiff and brittle structures usually do not perform well against seismic loads. Therefore. several structural and detail requirements are given on the lateral bracing structures and on the loadbearing joints. [1] 2. These forces are called seismic loads. thus. the building should be able to withstand vertical movements without loosing strength. The main emphasis in EC8 is on the security of human beings. in most cases. In this part.
masts and chimneys. This property is given by the 'action reduction factor' or 'behaviour factor' termed q. This factor lies in the range q = 1. Such analyses are listed below. 5. an acceleration value is normally used which has a return period of about 50 years. Lateral force (simplified modal response spectrum analysis) Multi-modal response spectrum analysis Power spectrum analysis Time-history analysis Frequency domain analysis 57 . 4. structures are classified in EC8 to several groups depending on their ability to deform and dissipate energy. 2006 Eurocode 8 part 1-3. The ground acceleration value of EC8. on a country by country basis in the relevant national application documents. The dimensioning of the structures is based on Eurocode 5 (EC5). This may depend on the seismicity. the nature of the building. This part handles the different building materials (concrete. Ay. These parts will not be considered in this report. The acceleration value is then Au. timber and masonry) and gives detailed structural requirements as well as detailing the specifications for buildings made of these materials. EC8 includes part 2 specific provisions for bridges. This depends on the seismicity of the area and is given. In addition to the above. 1. ag. NAD. Additionally. steel. National authorities may put restrictions on the building height or any other building property for seismic areas in their countries. The design philosophy for seismic events is that the building should withstand a so-called 'service earthquake' without serious movements or damage. the building should resist a socalled 'ultimate earthquake' without collapsing.3 Analysis methods for seismic design The seismic design of buildings may be carried out using several different methods of analysis. but shows the methods to determine the seismic loads with the acceleration value. the surrounding infrastructure or the foundation conditions. 2. In this case. on pages 158-164. EC8 does not contain these acceleration values. The part describing timber structures is in Chapter 8. part 4 specific provisions with respect to tanks. silos and pipelines. The return period of such earthquakes is around 475 years (EC8). Structures that have joints possessing plastic behaviour and energy dissipation can withstand much higher seismic events than structures with stiff and brittle joints. 2. part 5 specific provisions relevant to foundations. For this reason. retaining structures and geotechnical aspects and part 6 contains specific provisions for towers.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Specific rules for various materials and structures. is for the ultimate earthquake and its numerical value as well as its return period is given by the national authorities in the relevant national application documents as mentioned above.5–5 for timber structures. Portugal November. by the national authorities. it is necessary to know the peak ground acceleration value. ag in rock or firm ground (also noted as agR). This applies to all building materials. part 3 provisions for seismic strengthening and repair of existing structures. To determine the seismic load. but damages are allowed in this case. The ability of a structure to develop plastic strains and dissipate strain energy is central when determining its seismic performance (Ceccotti. 1989). 3.
Table 2. The plan configuration is compact. q reference decreased reference decreased * The fundamental period T0 should be less than 4×Tc and less than 2 seconds.in case of gradual setbacks preserving axial symmetry. It may be assumed that in most cases timber houses fulfil the requirements set on building regularity presented above. that building may be considered regular in elevation if the following conditions apply: • • • All lateral load-resisting systems run without interruption from the foundations to the top of the building or building part.In case of a single setback in the lower 15% of the total height of the building. the setback is not greater than 50% of the previous plan dimension.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. this does not exceed 6 % of the floor area. • According to EC8. 58 . The in-plane stiffness of the floors is sufficiently high in comparison to the shear walls. . without abrupt changes. only method 1.) exist. Considering setbacks on load bearing structures the following apply: . (Some additional rules for setbacks are given in EC8). Table 2. 'lateral force analysis'. According to EC8. If in plan set-backs (re-entrant corners etc. The deformation of the floor has a minor effect on the distribution of forces to the vertical stiffening elements.1 The significance of the building regularity (EC8) Building regularity Plan yes yes no no Elevation yes no yes no Simplification allowed Model planar planar spatial spatial Analysis lateral force * multi-modal lateral force * multi-modal Behaviour factor.1 gives the minimum requirements of the method of analysis. the setback at any floor is not greater than 20% of the previous plan dimension in the direction of the setback. 2006 EC8 permits the use of these analysis methods with certain restrictions. The slenderness of the building in plan is not higher than 4. a building may be considered regular in plan if the following conditions apply: • • • • The building is approximately symmetrical in plan in the two principal directions concerning lateral stiffness and mass distribution. can be carried out 'on hand' using the rules set by EC8. Portugal November.1. according to Table 2. these may be designed using the lateral force method. From these analysis methods. The other analysis methods are more complex and need special expertise. All lateral load resisting systems run without interruption from the foundations to the top of the building. Thus. which depends on the regularity of the building. Both the lateral stiffness and the mass of the individual storeys remain constant or reduce gradually. usually finite element methods are then needed which possess routines for dynamic analysis. from the base to the top.
International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. EC8 has a simple procedure: T0 = 0. of the building.3 Distribution of the base shear force in elevation (2) If the floor loads are equal in the different storeys. This is given by the equation: Fi = Fb z iWi ∑ j ziWi (3) Where Fi is the lateral load in storey i Fb is the base shear force zi is the distance of the floor from the ground Wi is the vertical load on the floor 59 .4) according to the ground conditions and the peak ground acceleration value. 2006 2. It should be noted that for a different country.4 Lateral force method The seismic design of a building starts with an evaluation of the regularity of the building in both layout and elevation compared with the requirements mentioned in section 3. the authorities may enforce values or parameters different from the ones given in EC8. The initial values are given. which are so-called boxed values. 2.3 . Portugal November. Fb = Sd (T0) W λ Where T0 is the fundamental period of the building Sd is the ordinate of the design spectrum W is the total weight of the building (see Chapter 6) λ is a correction factor.75 Where the building height is in metres and the time in seconds.4.0 otherwise.05 H0. the subsoil class (section 3. Usually timber residential buildings are regular in plan and in height.1 Base shear force The base shear force acts in both principal directions of the building.4. 2.85 if T0<TC or 1. 2. Generally regularity increases the seismic resistance of the building. The values given in this report are the ones recommended by EC8. according to the site seismicity. ag. the base shear force is distributed in a triangular manner so that higher forces are higher up.2 Fundamental period (1) To estimate the fundamental period. Such information is given in the national application documents. T0.4. having a value of 0.
10 0. Subsoil class B: Deposits of very dense sand.00 2. Average shear wave velocity vs <180m/s.4 0.2 1.6 0.50 2.25 1.5 ⎛ Tc ⎞ ⎜ ⎟ ⎟≥ βα q ⎜ ⎝ T0 ⎠ TT 2.b) (5. see section 3.0 1. 2006 2.8 0.2 1.3 60 . Subsoil class D: Deposits of loose-to-medium cohesionless soil.20 0.35 0.60 0.a ) (5.25 0.2 1.40 0.15 0.25 0.5 ⎞⎤ − 1⎟ S d = α S ⎢1 + 0 ⎜ ⎜ ⎟⎥ ⎠⎦ ⎣ Tb ⎝ q 2.4.15 0.5 1. Table 2.80 2.2 Parameters for the spectrum equations for the different subsoil classes (EC8). or any other soil profile not mentioned in the above classes.15 0.2 1.0 1.35 1. Average shear wave velocity vs 180-360 m/s. Average shear wave velocity vs 360-800 m/s.4 Design spectrum ⎡ T ⎛ 2.2 Subsoil class A: Rock or other rock-like formations. The national authorities will decide which type response spectrum will be used.2 Td [s] Type 1 recommended for large earthquakes A B C D • • • • • • • 1. Average shear wave velocity vs at least 800 m/s.30 1. Subsoil class E: A soil profile consisting of a surface alluvium layer with vs values of class C or D.50 0.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.10 0.00 E 1. Portugal November.05 0.00 2.00 2.00 1. Subsoil class C: Deep deposits of dense or medium dense sand.c) (5. the parameter values are as in the following table.15 1. of sensitive clays.25 0. gravel or very stiff clay.d ) According to the subsoil class. gravel or very stiff clay. For the behaviour factor q.5 Sd = α S q Sd = α S Sd = α S 2.5 ⎛ ⎜ c 2d q ⎜ ⎝ T0 ⎞ ⎟ ≥ βα ⎟ ⎠ jos T < Tb jos Tb < T0 < Tc jos Tc < T0 < Td jos Td < T0 (5. Subsoil class S Tb [s] Tc [s] Td [s] S Tb [s] Tc [s] Type 2 1.05 0. Subsoil class S1: Deposits containing a layer at least 10 m thick of soft clay/silts with high plasticity index (PI>40) and high water content Subsoil class S2: Deposits of liquefiable soils.05 0.20 0.
61 . power stations. ψ E . cultural buildings.2 schools.3 for live loads. ϕ is 0.i Qki (8) Where.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.6 Combining loads in seismic design (7) The design loads needed in seismic design consist of dead loads and seismic loads.i Qki (6) Gkj is the characteristic dead load and ψE.4 hospitals.i = ϕψ 2i ψ2i is the long-term value for variable loads 0.5 for all storeys except the top storey for which it is 1. ψE.0 residential and commercial buildings. However. The platform frame is very effective in this sense as the floor slab extends all the way through the external wall to the outer edge. if correlation exists then ϕ = 0. Also fire safety should be considered in seismic design.4. 2006 2.8 agricultural buildings).5 The mass in seismic design W = ∑ Gkj + ∑ψ E . The seismic action is cyclic and loading directions change constantly and simultaneously in horizontal and vertical directions. fire stations.1 Introduction The seismic design of timber houses is very similar to the design for other lateral loads such as wind loads. γ is the importance factor (γI = 1. γII = 1. Seismic design 3.8). EC8 does not give instructions on how to do this. Local fire safety rules and standards should be applied. The difference compared to wind loads is that the wind action is a pressure on the external wall whereas with seismic action the loads are connected with the masses (either dead or live) of the building and mainly directed to the floors. the structures should be firmly tied to each other and the floors or beams should not be able to slide from their supports. Therefore. The lateral loads are transferred to the foundations via the floor diaphragms and the shear walls.0 (no correlation between storey loads.2 for snow loads in Nordic countries and for altitudes above 1000 m and 0 for other countries 0 for wind loads (EC1 and EC5).iQki is the probable variable load during an earthquake.4. Gkj and Qki are the characteristic values of the dead and variable loads. Wind loads do not need to be considered.0 for storage loads (EC8) 2.i is the seismic combination coefficient of the quasi-permanent value of the live load. 3. The load-bearing wall supports the floor and the floor supports the upper edge of the lower floor. ϕ is 1. Portugal November. E d = ∑ Gkj + γFb + ∑ψ E . as fires may occur during a seismic event due to the cutting of electric wires or gas pipes. γIV = 0. γIII = 1. and 0.
The seismic load combinations should be considered when designing for the anchorage of the building for the following two cases: . V is the seismic shear load acting on the storey. d is the interstorey drift and h is the storey height.anchorage for sliding. Resistance The following condition should apply for all structures and connections E d = f {∑ Gkj .anchorage for overturning: upward tension at ends of shear walls. These will be considered in the following chapters. Portugal November. This will depend on the layout of the building and on the placement of the shear walls. The importance factor. Equilibrium The building should be stable during a seismic event. for residential buildings this has a value of 1. See section 3. The rules of EC8 for regularity were given in Section 2. Ductility The structures and the building as a whole should be adequately ductile. The anchorage for overturning is usually resolved either with bolted hold-downs or with nailed metal straps. 3. The material safety factor is γM = 1. Second order effects may not be considered if (P × d)/(V × h) < 0.0.3 for non-dissipative structures (Type A) and γM = 1. ∑ψ Ei Qki } ≤ Rd = R{ γM fk } (9) The design resistance of the structures is determined so that the material strength corresponds to the instantaneous load duration class.0 when the structure dissipates energy (Types B and C).3. γFb . ductility and equilibrium apply. the behaviour factor q was explained previously.10. The ductility should be considered in the design where it is taken into account as a load reducing factor. The building layout should be regular in elevation and in plan regarding mass and rigidity distributions. Also a shear panel crossing the storey level may be applied as an anchorage from store 62 . γ. 2006 The lateral stabilisation of the whole building should be designed so that the lateral loads from the different parts of the building are directed to the foundations.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. The lateral stability (or bracing) design of the building concerns the design of the floor diaphragms. was described in eq. .3 for the structure type classification.2 Ultimate limit state The safety of the structure for seismic events can be considered sufficient if the following conditions of resistance. base shear at the bottom of shear walls In a multi-storey building the anchorages should be considered at every storey level. Naturally the anchorage forces decrease at higher storey levels. the shear walls and the anchorage to the foundations. (8). where P is the storey (and above storey) gravity loads. The centre of the stiffness inertia and the centre of mass should coincide as closely as possible to avoid a torsion effect on the building.
5 and 5.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. These provisions will be met by all structures if : a) In doweled. To ensure the above classification of structures.1 mm.0 : Nailed wall panels with nailed diaphragms. trusses with nailed joints q = 4. the sheathing material is wood-based with a minimum thickness of 4 d. in this case the shear panel is in the outer edge of the wall as otherwise the floor would be in its way. the minimum thickness of the connected member is 10 d and the fastener diameter d does not exceed 12 mm. If the building is non-regular in elevation the above q-values should be reduced by 20 %. the most important rules and restrictions are given. Structures having medium capacity to dissipate energy such as: q = 2. q = 1. Hyperstatic portal frames with doweled and bolted joints Type C. but need not be lower than 1. connected with nails and bolts. two or three pinned joint arches. In case the building is stabilised for lateral loads with different structural types in the two main directions.5 for hyperstatic portal frames with doweled and bolted joints and q = 4. As stated before.0 Glued wall panels with glued diaphragms. q = 2. EC8 gives the behaviour factor q values for different structural types as below: Type A.5 . EC8 gives some additional rules and restrictions on the design of structures. 63 . 3. connected with nails and bolts. The anchorage for sliding is usually resolved with bolts or wedge anchors in the bottom plate of the shear wall in the lowest storey.5.0: Nailed wall panels with glued diaphragms. without more than 20 % reduction of their resistance.5. Structures having high capacity to dissipate energy such as: q = 3. the behaviour factor q for timber structures obtains the value between 1. which are specially applied for seismic actions. trusses with doweled and bolted joints. Portugal November. each one can be treated separately and a different behaviour factor for the two directions may be applied. In the following. In case the above requirements are not met. the nailing of the wall bottom plates should normally be strong enough to withstand the anchorage forces.0 for shear wall panels with nailed diaphragms.3 Special rules for timber structures For the different building materials. Type B.0 : Hyperstatic portal frames with doweled and bolted joints q = 5. the dissipative zones should be able to deform plastically for at least three fully reversed cycles at a static ductility ratio 4 for type B structures and 6 for type C structures. trusses joint with connectors. Structures having low capacity to dissipate energy such as: cantilevers. connected with nails and bolts. but the minimum thickness of 8 d and 3 d is assured for a) and b) respectively. bolted and nailed timber-to-timber and steel-to-timber joints. reduced q-values should be applied: q = 2. where the nail diameter does not exceed 3. in addition to normal structural design. mixed structures with timber frame and non-load-bearing infill. 2006 to storey. b) In shear walls and diaphragms. beams. In the upper storeys.
Portugal November. which are connected to top and bottom plates.4. with a load-bearing wall frame. and certain OSB. To this frame. however.5 Shear walls In a wooden house. the lateral stability of the house is most effectively provided by the use of shear walls with panel sheathing.2 g. The continuity of joists and headers should be ensured in places of diaphragm disturbances.4 Rules given by EC8 on the seismic design of floor diaphragms Floor diaphragms and shear walls may be designed in the same way as other lateral loads. 2 and 3.the minimum thickness of particleboard and fibreboard sheathing are at least 13 mm. shall be reduced by 75 %. The slenderness of the joists is restricted to h/b < 4. If the floor diaphragm is assumed to be perfectly rigid in-plan. the increase of 1. such as wind loads. by a factor of 1.2. such as panel corners. This should. • Particleboards. 3.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.5 for the nail spacing along the discontinuous panel edges may not be used. not result in a spacing less than the minimum spacing given in EC5. Fig. for example over the supporting shear walls. It should be emphasised. In seismic zones where ag > 0. Usually wooden load-bearing walls consist of vertical struts. 3. . equally spaced. the direction of the floor joists should not change.the minimum thickness of plywood plates is 9 mm. according to the procedures described in EC5 with certain exceptions. All edges of the sheathing panel should be supported and connected by blocking. that EC5 does state restrictions on the use of board materials for different humidity conditions: • Plywood boards may be used in service classes 1.1 Supporting and fastener spacing at the edges of sheathing panels according to EC8. may not be used.and fibreboards may be used in service classes 1 and 2.2 P(5) ). The shear forces may not necessarily be distributed evenly over the floor area ( EC5 5.the minimum density of particleboard panels is 650 kg/m3 . The following parts are different than given in EC5: • • • • • • • For floor diaphragms the increase in the fastener design capacity. a 64 . When sheets are staggered. and the in-plan position of the vertical shear walls should be considered. the spacing of the fasteners in areas of discontinuity. 3. 2006 The minimum requirements set on shear panels of shear walls are: .
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
sheathing panel is attached by nails or screws, on one or both sides of the frame. From a structural point of view, a shear wall may be regarded as a cantilever where a vertical load is located at the top plate. The sheathing transfers this vertical load to the foundations. The following text is based on references EC5 and STEP B13. Wood-based boards are often used as the sheathing panel, plywood or OSB. Gypsum boards may also be used. According to EC5, the shear capacity of the shear wall is based on the shear capacity of the fasteners as according to a lower bound plastic model. It is good practice to use the walls between attached dwellings and the walls next to corridors as shear walls; these may be sheathed on both sides and contain few interruptions from doors or windows.
y y
Top plate wall strut
Sheathing panel
Bottom plate (or sill plate)
Fig. 3.2 Basic shear wall and its static model. The wall struts are connected to the top and bottom plates with nails or with different types of metal connectors. These connections may be assumed to perform mechanically as a hinge. For this reason, the frame has to be braced with a sheathing panel, which is attached to the frame. The most loaded fasteners of the panel are located at the corners, where the displacement between the wooden frame and the panel is highest. The vertical struts are then designed only for the compression and tension forces along the edge of the shear wall. When panels of recommended thickness are connected to all the struts of the wooden frame with the usual strut spacing of K600 mm, the capacity to resist lateral loads is dependent on the fastener strength. Only in special cases, such as when using thin panels or having wide strut spacing, might the shear capacity of the panel or the shear buckling of the panel become decisive. According to EC5 the shear capacity of a shear wall is dependent on the fasteners capacity. When the fasteners are at equal spacing all over the panel and the panel width is at least h/4, the shear capacity may be calculated from (EC5 9.18): F v ,di = F f ,d bi s ci (10)
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
Where Ff,d is the design capacity of a single fastener, bi is the panel thickness, s is the fastener spacing along the panel edge (same on all edges) ci = 1 if bi ≥ b0 ci = bi / b0 if bi < b0 b0 = h/2, h is the height of panel The design capacity of the fastener, Ff,d , can be increased by a factor of 1.2 ( Ff,d = 1.2 Rd). The fastener spacing should be at the most 150 mm in the case of nails and 200 mm in the case of screws along the edges of the panel. In the central part of the panel, the spacing may be up to 300 mm, but never more than twice the spacing along the edges. The fasteners in the centre do not affect the shear capacity, except in reinforcing the panel against shear buckling. According to EC5 the shear buckling analysis does not need to be carried out if the following condition holds: bnet/t < 100 Where bnet is the free distance between struts and t is the panel thickness. So if the struts are spaced at K600, the buckling analysis need not be done because t > 5.95 mm. The total shear capacity of the shear wall may be calculated as the sum of the shear capacities of the panels (EC5 9.19): F v ,d = ∑ F v ,di
It is assumed here that the shear force is equally distributed along the fasteners connecting the panel to the wooden frame. The tension and compression struts at the sides of the shear wall have to be designed for the forces and the tension force has to be adequately anchored (EC5 9.20): Ft,d = Fc,d = Fv,d h/b (13)
The capacity of a shear wall composed of different panel elements may be calculated as the sum of the element capacities, even in the case where the panel material or fastener type is different. If a wall has the same panel sheathing and fasteners on both sides, the capacities of the two sheathings may be added together. If different panel materials are used, EC5 allows that 75% of the weaker panel capacity may be used, if the fastener strength-deformation curves are similar for the two panels, then only 50% of the weaker panel capacity may be utilised. If the wall contains a window, door or other opening, the shear capacities of these sections are omitted in the addition. The end struts of shear walls, as well as the bottom plate, should be anchored to the foundations to resist uplift forces (upwards) and sliding forces (horizontal). In a multi-storey house these anchoring forces should be considered from storey to storey as these accumulate towards the bottom storeys. Inner walls
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
The distribution of lateral loads to several shear walls depends on the rigidity of the floor and the rigidity of the shear wall. A rigid floor with flexible shear walls is one extreme case and a flexible floor with rigid shear walls is another extreme case. In the first case, the lateral force is distributed to the shear walls depending on their relative levels of rigidity. In a case where the floor is supported by three shear walls of equal rigidity, each of these walls carries a third of the lateral load. However, if the inner wall is not located at the centre, a torsion component is also developed. In the other extreme case, the floor may be regarded as a continuous multi-span beam over the supporting shear walls or two non-continuous single-span beams extending between two shear walls. To be on the safe side, it may be good practice to design the outer shear walls assuming a single floor span supporting conditions and the inner shear walls assuming continuity of the floor. The assumption of a perfectly rigid floor should only be used if the floor plan dimension ratio is close to one. Calculated capacities of shear walls with plywood sheathings
Shear capacity of different sheathing panels in seismic design 14 Fasteners screw 3.5 (3.9x29 Gyprocille) K100 mm 12 15 mm spruce plywood 10 9 mm spruce plywood 12 mm spruce plywood 21 mm Kerto-Q LVL 27 mm Kerto-Q LVL
Gypsum board (Gyproc GN13)
0 1 2 3 4 5 6
Fig. 3.3 A comparison of shear capacity per wall length of different sheathing panels. The following table contains calculated shear capacities, which may be used in the seismic design of timber buildings braced with shear walls. Calculated cases with several different plywood and LVL sheathing panels and different fasteners are given. These design values were calculated for instantaneous load duration and using a normal material safety factor, as EC8 assumes an energy dissipative structure.
3.6 The anchorage of the building
In order to transfer the lateral loads to the foundations, the building has to be anchored to the storey below and then on to the foundations. Anchoring is normally required at the ends of shear walls to account for the uplift forces (due to overturning when the building's own weight does not compensate for the effects of the lateral load) and at the bottom plate to account for the sliding (slip from base shear), see Fig. 3.4. The uplift and sliding forces are anchored independently of each other with special connectors.
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
Fig. 3.4 The two anchoring cases: sliding caused by base shear and uplift caused by overturning (APA 1997) Figure 3.7 demonstrates the anchoring of a shear wall of a single storey house and Figure 3.9 the anchoring of a multi-storey house. The basic principle is as follows. - To withstand overturning, both sides of a shear wall are anchored to the storeys below and to the foundations to account for uplift forces. The connection is fastened to the frame strut and this transfers the tensile forces downward to the struts below or to the foundations. The connection cannot be attached only to the panelling or only to the frame plate. The fastener may be a holddown bolt with metal end-plates or a metal strap that is nailed to the frame. In the case of a loadbearing wall, the self-weight may be subtracted from the uplift force. - To prevent sliding, shear anchor bolts are usually used to connect the bottom plate (sill plate) to the foundations. The spacing of these bolts will result from the calculations. For upper storeys, the nailing of the bottom plate to the flooring or to the wall section below is usually sufficient to account for the shear forces at these upper levels. The shear force acting on a single fastener is:
F = Fv ,d s l
Where Fv,d is the shear force (lateral load) acting on a shear wall s is the spacing and l is the length of the shear wall
. which returns a single base shear value acting on the building. Dean. Also. J. 3. References [1] [2] [3] [4] Seismic design of timber structures according to Eurocode 8. local authorities should be better informed on the performance of timber houses. in areas of high seismic activity. Portugal November. Seismic design expertise should also be widely known in countries. Usually. 4. International Timber Engineering Conference. both in plan and in elevation. However. wood design concepts. In this way. (available from the author as a pdf file). When the seismic load is calculated. A. (1988) Practical design of timber structures to resist earthquakes. (2000) Seismic behaviour of timber buildings. Seattle USA Buchanan. The good seismic performance of timber buildings should be used in the marketing aimed at seismic areas of the world. 69 . COST E5 Workshop on Seismic behaviour of Timber Structures. which do not have earthquakes. Espoo Finland 2002. the lateral loads should be taken into account early in the design process when planning the layout and the frame. (1997a) Earthquake safeguards. Introduction to lateral design. In the case of multi-storey timber houses. buildings are made of concrete or masonry and these can be very dangerous if not properly designed for seismic actions. A. the local requirements for houses can be better fulfilled. some restrictions on the detailing of floor diaphragms and shear walls are given in EC8. The seismic design may then be carried out using the simplified modal response spectrum analysis. Conclusion Timber houses are usually regular. 5. Therefore. but which do have a timber house exporting industry. the bracing is designed according to EC5.5 The anchoring of uplift forces is handled with metal straps or hold-down bolts (SECBC 1997). EC8 gives the rules on how this base shear is calculated. Ceccotti. VTT Research notes.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. introduction. 1988: 813-822. The bracing of the building is designed in both principal directions against this base shear load. the seismic loads are about twice the magnitude of the wind loads in high seismic zones. Seattle. APA. 2006 Fig. September 28-29 2000 Venice Italy.
[11] SECBC (1997) Updated seismic design of buildings. Draft No 4 December 2001 CEN.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. TC 250 of CEN. Belgium. Centrum Hout 1995 70 . (1989) Structural Behaviour of Timber Constructions in Seismic Zones. Basis of design. Centrum Hout 1995 [13] STEP C17 Timber connections under seismic actions. September 2829 2000 Venice Italy. Belgium. A. Design of structures for earthquake resistance.Univ. Brussels. June 1998 UK. material properties. Part I: General rules. (1998) Seismic design considerations on the multi-storey woodframe structures. seismic actions and rules for buildings. Karacabeyli. A. Ario Ceccotti. STEP 1 Timber Engineering. STEP 1 Timber Engineering. EUROCODE 5 (1995) ENV 1995-1-1. Basis of design.. of the relevant CEC DG III . EUROCODE 8 (2001) prEN 1998-1. editor. 2006 [5] Ceccotti. Wood Frame Committee. Portugal November. Florence. RILEM State-of-the-Art Report. Cost E5 workshop on constructional aspects of multi-storey timber buildings. structural components and joints. [6] [7] [8] [9] [10] RILEM TC 109 TSA (1994) Timber structures in seismic regions.1999) Karacabeyli. Ceccotti. Thomas Alsmarker. [12] STEP B13 Diaphragms and shear walls. material properties. Brussels. Design of timber structures..06. European prestandard. Proc. structural components and joints. COST E5 Workshop on Seismic behaviour of Timber Structures. (Version 29. (2000) Performance of North American platform frame wood construction in earthquakes. Structural Engineering Consultants of British Columbia. of Florence Workshop. Italy. E. E. Material and Structures 27: 157-184.
fournely@cust. In fact. design detailing is always an essential stage. of the AFPS guide document on the design detailing for earthquake engineering. Portugal November. On one hand it is an interesting event. and also experts of control offices and universities. In France. The timber structure chapter is the result of a working group including specialists of design in factories. appropriate solutions are proposed and different options for principal joints are defined. new applications are possible by the integration of new knowledge. if training is necessary to use correctly new standards.univ-bpclermont. The timber structure chapter is organised around the different typologies of usual timber structures: glued laminated timber frames. new fabrication or erecting technique. more particularly in the field of timber structures. It is a way of innovation. France thierry. France eric. The arrival of a new code is always a strategic period for the industries. but progress and development can also appear with the application and guidance for this application of the standards. This AFPS document (AFPS: French Association for Earthquake Engineering) proposes examples of solutions for the different joints and specific elements with. information or guides French government by the way of its different Departments and specific directions proposes various actions in order to provide answers to civil engineering industry companies. Bureau Veritas Technical Department La Défense.technical documents. The composition of this WG is indicated in references. it is also a strategic time to begin new applications in accordance with the scope of the code. traditional carpentry and industrial timber trusses with punched metal plate fasteners. for building structures different actions has been or are realized. the application for timber structures.fr Thierry Lamadon Princ-Ing.lamadon@fr. The aim of this presentation is an illustration of the contents. On the other hand. Civil engineering laboratory – CUST Blaise Pascal University Clermont-Ferrand. In AFPS guide. and the organisation. For each kind of structure. This aspect is much more important for timber structures due to the fact that dissipative zones are principally located in joints. 2006 Design detailing for earthquake engineering Application for timber structures Eric Fournely Dr. 71 . Introduction of Eurocodes and more specifically Eurocode 8 are good examples of these items. Figure 1 illustrates these different actions for Eurocode 8. a global approach is discussed and justified. Design detailing is only evocated in standards and only few documents are available to present such detailing. simplified approaches and illustrated solutions are necessary in order to give examples and guidelines. engineering design.com 1 Introduction Innovation can be provide by new material. .training actions with pedagogic and simplified documents and course propositions. The main aims of these actions cover: .bureauveritas.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. for professional who already work in the domain. and also for other professional who want to work in this domain. in our case. In earthquake design process. multi storey buildings with traditional bracing or shear walls.
International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Portique en pignon Gable with portal frame Portique Portal frame Poutre au vent Roof trusses Action sismique Longitudinal dans le plan earthquake action longitudinal du bâtiment Action sismique transversal dans le plan earthquake action transversal du bâtiment Poutre au vent Longitudinal longitudinale roof trusses Longitudinal bracing Palé ée de stabilité é Pal stabilit Pignon en ossature Braced gable b) Bâtiment à 1 niveau avec portiques courants et pignons en ossature traditionnelle a) Bâtiment à 1 niveau avec portiques courants et portiques en pignon Figure 2: Longitudinal and transversal stability of single storey portal frame building 72 . Figure 1: Overview of action performed or started in the field of earthquake engineering and more particularly for timber structures 2 Glulam timber frames For classical timber portal frames. the necessity of a roof diaphragm is discussed. erecting and transportation). Compatibility of displacements is correlated to roof covering and cladding stiffness. The next paragraphs of this paper present an illustration of these propositions. If diaphragm does not exist regularity criteria based on diaphragm effect are not available. or attachment. bracing system force transfer is described. Links with foundation and masonry chapters are proposed. 2006 more than 60 original configurations of usual joints are presented. or for braced gables. In the other direction. for example by juxtaposition of elementary modulus likes those presented in figure 2. Glued laminated timber portal frames allow covering wide areas. In this case appears cantilever joints (facilities for handing. Figure 2 gives an overview of these approaches. ductility. Portugal November.
Portugal November. Multi bay portal frames with cantilever beams Figure 3: Details of a ridge joint. column base joint and cantilever beam for different portal frames 73 . 2006 Figure 3 presents an example of a ridge joint. a column base and a cantilever beam joint.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. The development of tension perpendicular to the grain in joint zone is particularly studied. These joints are presented as illustrations.
in EN1995-1. In order to allow larger values of behaviour factor than DCM ones.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. In EC5-DAN French version. These proposals are reported in table 1. and diaphragm flange continuity are studied and developed. Without more information. static ductility classes were mentioned and. Figure 4 gives examples of global configurations of multi-storey timber buildings and some detailing of joints. A particular attention is oriented on the single 74 . The second objective in the choice of joint configurations is to obtain ductility in main joints. storey continuity. end and edge distances corresponding to XP EC5 values.1. This objective induces also consequences on the global framing solution such as portal with V-shaped frame geometry (with cogging joints for example) and without knee mechanical joints. obviously. anchorage. they must be applied for all kinds of timber joints. Portugal November. the AFPS group chose XP EC5 values. Theoretical elements on ductility of multi-doweled joint do not still exist. edge and end distances are presented in this paragraph but. French recommendations limit q factor values to those of DCL or DCM class for dowel type jointed structures. 2006 The first objective is to limit the existence and the intensity of this tension perpendicular to the grain. The ductility of mechanical timber joints is not so well known. The joint configurations are proposed in accordance with this aim. The proposal in this aim is to choose spacing. The base of the theory of design of these joints is essentially based on resistance. These conditions on spacing. In fact. the care must be taken in performing experimentations in accordance with EN12512. these values have been reduced (specifically a1) in comparison with previous version. Table 1: Minimal spacing and edge distances for dowel type fasteners [XP EC5] 3 Multi storey buildings with traditional bracing or shear walls Steel and timber bracing. to propose design rules.
1 proposes different methods (A and B methods) for the validation of shear walls. and the ductility. Figure 4: Examples of multi storey timber building with bracing. These criteria are notably discriminating in the case of important openings in floors. Portugal November. EN19951. and basements. are essential points in the reliability of the whole structure undergoing seismic action. the observed ductility is higher than this obtained with over constructive techniques. between shear walls. the diameter and spacing of nails or screws. In these conditions. with the impact of the different openings.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. particularly developed for the shear walls. 2006 intersection of member axis and the slenderness ratio of the bracing. figure 5). The beam mesh of the floor has to respect bending criteria. Generally. but. Criteria of largest rigidity (compare to the lateral one of the vertical structure) and of over strength according to capacity design have to be controlled. the ductility of the internal joints of the floor is not required. Nevertheless. the floor must assume diaphragm function. details diagonal and beam floor joints The other points. and much more information is available on cyclic or seismic experimental results (cf. with screwed or nailed panels. 75 . are the minimal thickness and density of the panels. the analyses of effective continuities.
Portugal November.e. As mentioned in EN 1998-1. details of framed joints and anchorage.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 76 .1). Results of timber structures with framed joints in EN12512 loading conditions could allow more realistic evaluation of behaviour factor and also an other ductility classification including the impact of γm factor. 2006 Figure 5: Examples of multi storey timber building with shear walls. "Annex" apparatus are also particularly studied in order to guaranty stability in plane and out of plane of the shape of the timber framing. EN 1998-1.1.5 correspond to a careful position. constitution. Figure 6 gives illustrations of 3D element distribution for a timber framing on an L-shape building. the over strength factor for capacity design is used to justify the widest dimensions of the section undergoing shear stress for example (i. anchorages and generic behaviour 4 Traditional carpentry For traditional carpentry. Nevertheless. Without any more information. q=1. anchorage with concrete or masonry walls is widely discussed. lv value). joints.1 does not propose clear values of q factor (carpentry joints with force transfer by contact are not mentioned in table 8.
Different solutions of horizontal trussed frames are studied and proposed in correlation with dimensions of the building and the seismic activity of the site. Different values of q factors are proposed for industrial trusses combined with timber walls or with masonry. ductile enough with an effective control of the anchorage.3] 77 .International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. or configuration. < 0. industrial trusses out of plane elements spacing of trusses french name or abbr. Anchoring flat positions are indicated and linked with masonry chapter. framed joints and anchorage for ancient timber frames 5 Industrial timber trusses with punched metal plate fasteners Out of plane low strength and rigidity of concrete or masonry walls is exhibited. Figure 6 illustrates also these aspects. Several solutions for the bracing in the cases of singularities in the roofing are proposed. The mass of the masonry walls and the quite different companies involved in the erecting (and in design) of the building induce to take care in effective reliability of the whole structure. Dimensions of out of plane trusses elements and quality of fasteners are essential for the quality and the reliability of industrial trussed systems. stiff. Bracing in the plane of the roofing is widely commented.70 à 0. Figure 7 illustrates these joining and composite behaviour. Table 2 reports the values of this dimensions according to states of arts and French prescription document DTU 31. Quite different forces are transferred by joints with these joining types.3.92 m AFA 25x100 36x96 FAA & FA 25x60 25x72 CVS 25x72 25x72 ENT 36x72 36x72 ADA 25x100 36x96 FE & FD 25x60 25x72 minimal dimensions (H=12%) of out of plane elements mm² Table 2: Minimal sections of industrial trusses bracing [DTU31. The third type is more specially design for timber walls. Portugal November. Type and diameter of nails must be adapted (thickness of trusses elements) and mechanical angle bracket have to be thick. Three types of joining between trusses and walls are discussed and justified. 2006 Figure 6: Global configuration.70 m 0.
Usual anchors are synthetically presented. This figure resumes usual configurations of glued laminated timber and solid wood joints with concrete or masonry. 78 .International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. This is one of the principal points to be check in the design documents and in realization. 2006 Figure 7: Horizontal and roofing plane bracing for light industrial timber trusses 6 Joining and anchorage with masonry and foundation Figure 8 presents other joints. more specially oriented to interface between timber and basements. A lot work still to be made in order to dispose a wide panel of possible solutions with guarantied mechanical characteristics and reliability in seismic conditions. but for mechanical aspects more precise indications or more engineering investigations must be performed. Portugal November. Attestation of CE mark gives indication for minimal durability.
the possibility will be given to companies to have calibrated solutions for the different seismic zone map. Recommended values of q factors proposed to the different timber structures in this guide document are limited to DCL or DCM ones or the lowest of DCH ones (shear walls q=3).7 with the new Seismic Hazard Map (proposed by French government). simplified rules according to EN 1998 1-1 must be defined and drawn up. rules and guidelines is also induced by the increasing of national territory including earthquake engineering approach for design. Portugal November. In order to complete the work made under the umbrella of this action. figure 9). 2006 Figure 8: joining with masonry and concrete walls or columns 7 Conclusion This paper is only a sampling of the result of the work made in the WG of AFPS. This work is begun for individual houses and for usual R+2 and R+3 buildings. The necessity of these documents. 79 . Seismic area of application of the earthquake engineering should be multiplied by 3. By this way.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. and reached more than 60% of France area (cf.
Pr NF EN 1998-1/NA. tomes 1 et 2. G. 8 References . Structures en bois aux états limites. September 2002.XP ENV 1995-1. T. Cabaton GTM Fargeot.com Figure 9: Comparison between seismic map and timber construction activity in France A first step of innovation for companies and engineering design offices can be provided by simplified rules and technical guidelines in order to have basic solutions. Cours élémentaire de dessin de charpente 80 .1 Conception et calcul des structures en bois – partie 1. Calvy Direct Ingénierie. corresponding to wide possibilities of new development. CIB W18. J-P. Portugal November. Projet de l'annexe nationale française de l'Eurocode 8. Lemaire SOCOTEC (and CTICM) 2005 . méthodes d'essais : essais cycliques d'assemblages realisés par organes métalliques.STEP/EUROFORTECH. . .document d'application nationale (EC5-DAN) . (DTU 31. V. Design of timber structures in seismic zones according to Eurocode 8. Toratti. . Kyoto. 2006 Seismic Hazard Map Business-in-europe.Background documents of WG timber structures AFPS M. pp 231302. Th. Japan. final draft .International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Fournely. Presses des ponts et chaussées.A. Fournely C/U/S/T/. béton. 2006 .NF P 21-205 Charpentes en bois assemblées par connecteurs métalliques ou goussets – partie 1 Règles de mise en œuvre. is to identify clearly the cyclic and dissipative behaviour (EN 12512) of different typologies of joints in order to estimate the global ductility of the building. Biger expert. juin 1996. Lamadon et al. Betton C/U/S/T/.ouvrages en bois. de Zutter MITEK.EN 1998 1-1 Design of structures for earthquake resistance – part 1 : General rules. E.Encyclopédie Delafosse. Lamadon Bureau Veritas. seismic actions and rules for buildings. E. Editions Eyrolles. V. A second step. . janvier 2006.M.NF EN 12512 Structures en bois. bois et maçonnerie. Ceccotti. Gianquinto. Guide des dispositions constructives parasismiques des ouvrages en acier. L.1 Règles générales et règles pour les bâtiments. Th.3) .
a region of the northest of Italy particularly rich of wood) and funded by the Autonomous Province of Trento. of Structural Engineering. Shizuoka University.it CNR-IVALSA Via Biasi. Prof. and they're gaining increasing shares in the residential market. Member of Eurocode 5 (Timber Structures) and Eurocode 8 (Structures in Seismic Areas) Civil Engineer. 81 . • the system shows a good ductility and good overall dissipating properties in dependence of the Fig. therefore they allow the use of medium-low grades of homegrown sawn wood. where Fiemme is a part of Trentino.ceccotti@ivalsa. Japan. Building Research Institute and Center for Better Living. Portugal November. 1. The test were conducted at the NIED Tsukuba Shaking Table facility in June and July of 2006. It is the first time that such a test have been ever conducted on this type of structures. Research Associate in Timber Engineering and Seismic Behaviour of Timber Structures for the Trees and Timber Institute of the National Research Council of Italy (CNR-IVALSA) and University of Florence.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Their unquestionable success is due to several reasons among which we may take into account the following: • the cross lamination method gives a material with high stability and good overall mechanical properties. Introduction XLam structures (where XLam stands for cross-laminated timber boards) are becoming in the last years a "standard" in Europe among the timber-based constructive systems. In this paper results from shaking table tests on a full-scale 3 storied XLam building are presented. 2006 Seismic Behaviour of Multi-Storey XLam Buildings Ario Ceccotti Professor Director ario. 1 Four stories XLam building connection layout. good thermal insulation. Summary The Trees and Timber Institute of the National Research Council of Italy (CNR-IVALSA) in collaboration with National Institute for Earth Science and Disaster Pevention (NIED). initiated in 2005 a research program on the seismic resistance of Multi-Storey XLam Buildings.it Maurizio Follesa Civil Engineer Research Associate follesa@timberengineering. and a fairly good behaviour in case of earthquake or fire.cnr. Italy. even for multi-storey buildings. University of Venice. This project is undertaken in cooperation with many international research institutions and deals with many aspects of this new constructive system in order to enhance the use of wood and of multistorey wooden buildings in Italy and Europe. 75 38010 San Michele all'Adige (TN) Civil Engineer. • XLam panels are extremely strong and stiff whatever is the timber quality. This is a part of a more comprehensive program named SOFIE Project (in which SOFIE stands for the Fiemme house constructive system. Director of the Trees and Timber Institute of the National Research Council of Italy (CNRIVALSA).
design. up to now European Standards are not yet supporting this constructive system with an harmonised set of product. erection. less keen to "lightweight" timber buildings systems and more familiar with massive masonry buildings. Therefore more research is needed in order to obtain performance data to support the acceptance of seismic provisions of multi-storey XLam buildings for Eurocode 8 and the Seismic Code of Italy. in-plane monotonic and cyclic tests on full-scale wall specimen with different opening 82 . The shaking table tests described in this paper are the last phase of an experimental program on the seismic behaviour of this type of buildings started in Italy in 2005 which comprises also tests on connections.g. very few research programs have been carried out up to now and no one of them including a shaking table test on a full-scale specimen. due to the fast growing rate of diffusion of this system. 2006 • the XLam system allows both for single unit housing and multi-storey buildings. Portugal November. • the construction process is very quick and possible even for non-highly-skilled manpower. However. the height limits for wooden buildings given in the Italian Seismic Code).International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Especially considering the seismic behaviour of this type of structures. and most of all. inspection and maintenance rules which may summon up the know-how and the design and erection procedures already developed. This "lack" of research results is reflected by the "lack" of design rules in European standards for the seismic design of structures. The construction process is very quick and possible even for non specialized craftsmen. the q value in Eurocode 8. a b c d Fig. and the few one which may be found are too much conservative (e. • the XLam system is more appealing to a large part of European public. 2 Construction phases of the 3 storey XLam building tested in NIED Tsukuba Shaking Table in June and July of 2006.
2m openings for windows.25m opening in the middle at each floor. The 2nd and 3rd level outer walls of the building had all two 1. The house was connected to the steel base with commercial type holdown anchors and steel angles.1 500 ton 14. The panels were made with spruce coming from the woods of the Val di Fiemme. The two floors were also made with XLam panels of 142 mm of thickness and the roof panels had a thickness of 85 mm. 3 The test house on the shaking table The building was constituted of 4 outer walls which were made of XLam panels of 85 mm of thickness and one inner wall parallel to the E-W direction (the direction of shaking) of the same thickness with a 2. The basic specifications of the table are listed in Table 1 Table 1 Basic specification of the shaking table Payload Size Shaking Direction Maximum Acceleration Maximum Velocity Maximum Displacement 2.P. 2006 and connection layouts and a pseudo-dynamic test on a one-storey specimen with the same dimensions of the one described in this paper. as illustrated in Figures 2 and 3. The house was built and dismounted after the tests directly on the shaking table by a team of Italian and Japanese carpenters. Portugal November.5m x 15m X Horizontal 500cm/s2 (500 ton) 940cm/s2 (200 ton) 90cm/s ±22 cm The test specimen was a 3 stories house of about 7 m x 7 m in plan and 10 m of total height with a pitched roof.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Sandhaas. The final purpose of this research project is to develop a reliable analytical model capable to predict the non-linear behaviour of this kind of buildings under real earthquakes excitations in order to provide the European and Italian codes with a quantitative basis for the rational calculations of the seismic behaviour of XLam buildings. Lauriola and C. The results of this preliminary tests are described in paper "Quasi-Static and Pseudo-Dynamic Tests on XLam Walls and Buildings" by M. Fig. 2.1x1. At the Description of Test Specimen 1003 83 . Shaking table tests The tests were conducted on the 14. The connection to the shaking table was made by means of a steel frame made of H 300x300x10x15 profiles according to JIS G 3192: 2000 which was itself connected to the shaking table by means of ø50 bolts.4m x 2. Trentino in the North East of Italy and were delivered from Italy to Japan.5m x 15m NIED Shaking Table in Tsukuba.
5. 4 Three different configurations in which the building have been tested.4m).0m. while the two external walls parallel to the shaking direction (E-W direction) had each one a door opening whose width was varying in the three subsequent phases of the test (see Fig.2 and 2.0m. two of them placed above and below 84 . system.8 Class M16 anchor bolts and with ø4 annular ringed nails to the walls) placed at wall endings and at door openings and steel angles (BMF 90x48x3x116 connected to the basement by means of 8. Configuration C is asymmetric as the width of the openings in the two external walls parallel to the shaking direction were 2.2m to 4. Also the floor panels had the same limitations and were connected together with overlapping joints made with ø10x180 selfdrilling screws.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 5 Vertical joint between wall panels and horizontal joint between of ductility of the entire floor panels and self-drilling screws used to connect panels.30m. The inter-storey connection between walls and floor was made again with holdown anchors at corners (Simpson StrongTie HTT16.2m openings each.4 and 4.1x1. Fig. The connection to the steel base was made by means of commercial type holdown anchors (Simpson StrongTie HTT22 connected to the basement by means of 8. The choice of making vertical screwed joints between wall panels was not only taken for transportation reasons but was also a defined design issue in order to achieve the desired level Fig. 4) from 1. 2006 1st floor the two external wall perpendicular to the shaking direction (N-S direction) had again two 1. Therefore each wall is constituted of 3 panels connected together with a vertical joint made with LVL stripes and ø8x80 self-drilling screws. Portugal November. While Configuration A and B are symmetric (the door opening widths in the two external walls parallel to the shaking direction were respectively 1. For transportation reasons each floor and wall panel had not to be wider than 2. The same screws were used in the vertical joints between perpendicular walls and to connect floor panels to the walls below as showed in Fig.8 Class M12 anchor bolts and with ø4 annular ringed nails to the walls) distributed along the length of each wall.
Fig. 2. The base shear force was therefore calculated according to the following equation: Fb (T1 ) = S d (T1 ) × W (1) Where Sd (T1) is is the ordinate of the design spectrum at period T1 and W is the total mass of the building. 1005 85 . or very stiff clay) q is the behaviour factor which was taken equal to 1. deposits of very dense sand.2 Design of the test building The test building was designed according to Eurocode 8.8 Class M16 anchor bolts and with ø4 annular ringed nails to the walls) and steel angles distributed along the length of each wall (BMF 90x48x3x116 connected to the floor and to the walls with ø4 annular ringed nails). 2006 the floor panel and connected together by means of 8.5 q (2) where: ag is the design ground acceleration corresponding to the seismic zone. The building was designed following the Capacity Design Method. e. 6 Holdown anchors and steel angles used to connect wall panels to the steel base and to the 1st floor.20 s.g. From the outcomes of the preliminary tests made in Italy the period T1 of the building was estimated to be 0. e.35g. Thus ensuing this strategy the vertical joints between perpendicular walls and the horizontal joints between floor panels were over-designed and the building was designed in order to reach the energy dissipation first in the vertical joints between wall panels.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. then in the horizontal connection between walls and floors (steel angles and screws) and last in the holdown connection. corresponding to the most hazardous value of the Italian territory S is the soil factor (taken equal to 1.25 accounting for type B soil.g. therefore the ordinate of the design spectrum is S d (T1 ) = a g × S × 2. gravel. According to the Italian Seismic Building Code is taken equal to 0. Portugal November. as in the definition given in Eurocode 8 the design method in which elements of the structural system are chosen and suitably designed and detailed for energy dissipation under severe deformations while all other structural elements are provided with sufficient strength so that the chosen means of energy dissipation can be maintained.
Following the same strategy the holdown connections at 2nd floor had 9 nails and 5 nails at 3rd floor. 2.0 for snow loads.3 for live loads and 0. placed randomly in the upper side of the holdown connection in order to reach a higher dissipation of energy due to the plasticization of the steel in the holdown anchor.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. For instance the holdown connection to the basement was calculated with 12 nails. 8. 7 Holdown connection to the steel base with 12 nails. therefore no additional weights on the roof have been placed. Summarizing the total weight considering the test house. 8 Weight distribution at 1st and 2nd floor 86 . Portugal November. According to the Italian National Building Code the characteristic values of live loads on each floor of a residential building should be 2. thicknesses and weights as showed in Fig. not more not less. Fig. the steel base and the additional loads is the following: Table 2 Load distribution at each floor Floor Steel base 1st 2nd Roof Total Self weight [kN] 35 60 60 45 200 Additional [kN] 150 150 0 300 Total [kN] 210 210 45 500 Additional weights have been provided by means of steel bases of different sizes.3 Additional weights Additional weights have been added to each floor to account for the weight of finishing and insulating material typically used in this construction system and to take into account the presence of the 30% of live loads in seismic load combination as prescribed by European and Italian seismic code.0 kN/m2. The combination factor for earthquake action is 0. This strategy was pursued even in each single connection in order to reach for the whole structure the maximum ductility and energy dissipation as possible. 2006 Fig. starting from the results of tests on connections and in-plane monotonic and cyclic tests with different connection layouts preliminary held in Italy.
On these towers 8 wire type displacement transducers (2 per level) and 6 laser transducers (2 per floor) were placed plus 2 relative displacement transducers to measure slip between wal panels and steel base. Portugal November. 2 in the N-S direction and vertical direction (6 per floor) plus 2 were placed on the roof in the E-W and N-S direction.5g) as showed in Fig. 2 at each corner at the ground level in correspondence of the holdown anchors. 9 Different type of instruments: accelerometers (a). These were made by means of thick sheets of plywood fixed to the ceiling and to the floor with inductance type displacements transducers. relative displacement transducer to measure inter-storey drift (b). For each phase the building was tested with 3 different earthquakes (Kobe. The same displacements were measured by means of other 12 instruments (4 per floor) made with aluminium bars and strain gages. B and C with three different opening layouts in the two external wall parallel to the shaking direction.15g and 0.. El Centro and Nocera Umbra) at 2 growing levels of PGA (0. named respectively Phase A. 3). load cells (d). To measure relative displacements of each floor relatively to the base along the shaking direction two steel towers were placed on the East side of the test specimen and fixed to the shaking table (see Fig. 14 relative displacement transducers were applied (6 at the ground level and 4 at 2nd and 3rd floor) and also 12 relative displacement transducers were placed to measure the slip between wall panels and floor panels.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. wire type displacement transducer (c). A total of 20 accelerometers were placed. At each floor 2 respectively in the E-W direction (direction of shaking). Some of them replicated the same measurements but with a different device. 2. To measure the inter-storey drift 15 instruments were placed (5 per floor) measuring displacements in both horizontal direction. To measure uplift.4 Instrumentation The test specimen have been instrumented with 97 instruments. three different phases of the test have been planned. a b c d Fig.5 Test Procedure In order to study the behaviour of this construction system with reference to the length of resisting walls and the torsional behaviour in case of asymmetric layout of stiffnesses. Finally 8 load cells were inserted between the steel base plate and the nuts. 1007 87 . 3 and 10. 2006 2.
90 Fig.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. the specimen have been examined for evidence of damage or deterioration. Each time before and after each level of excitation the model dynamic parameters have been measured using both a step impulse and a white noise in order to catch the changes of the natural frequency of the model and the ratio of damping so to evaluate the extent of how the structural stiffness would eventually decrease.340 2.468 0.94 6.868 2.94 6.935 1.935 6. 10 Opening layout of 1st storey walls for Phase A.10 0.15g and 0.935 1. 88 2. and any changes have been documented and photographed.5g or more.340 2.868 2.340 2.340 2. 2006 PHASE A CONFIGURATION FIRST FLOOR PLAN PHASE B CONFIGURATION FIRST FLOOR PLAN PHASE C CONFIGURATION FIRST FLOOR PLAN 2. Finally configuration C of the building have been tested also with Kobe and Nocera Umbra with growing level of excitations up to the reaching of the "near collapse" status.90 1.0 13. In each shaking test marked with bold character.255 4.499 All the three earthquake records have been scaled with respect to the PGA and for each phase of the test have been inputted at 0.7 0.868 2.94 6.255 2. Table 3 Details of chosen ground motion records.340 2.35g.468 2.313 0.935 1.255 2.255 1.10 2. 14 of which with a peak ground acceleration of 0.935 2. Before and after any intervention again the dynamic parameters have been measured in order to catch the change of the natural frequency. whose PGA is 0.340 1.935 6.340 2.868 2.10 2. Umbria Date 1995/01/16 1940/05/19 1997/09/27 Station JMA Imperial Valley Nocera Component N-S N-S E-W Duration (s) PGA (g) 48. A total of 26 shake table motions were inputted.90 1.340 2.340 0.340 2.20 2. Besides an intervention of repair have been foreseen by means of replacing screws in new positions in the vertical joints between wall panels and retightening of anchor bolts of holdown and steel angle connections. The details of the earthquake records chosen as input of seismic excitation are listed in table 3.90 1.50g respectively corresponding to hazard level of Zone 3 and more than the Zone1 of the Italian Seismic Code.10 0. In case of more severe damage (damage to holdown anchors and or steel angle) the intervention of repair was made by moving connectors in new positions.821 0. Portugal November.90 0. B and C respectively.000 6.90 1. Record Name Kobe El Centro Nocera Umbra Country Japan California Italy.20 1. The shaking have been operated unidirectionally in each phase of the test. Summarizing the steps of the whole test have been the following.255 DIRECTION OF SHAKING DIRECTION OF SHAKING DIRECTION OF SHAKING EXTERNAL WALL THICKNESS 85 mm INNER WALL THICKNESS 85 mm EXTERNAL WALL THICKNESS 85 mm INNER WALL THICKNESS 85 mm EXTERNAL WALL THICKNESS 85 mm INNER WALL THICKNESS 85 mm 2.340 2.0 40.10 2.10 0.340 W N E S .255 2.
2 1.2 1.255 06070305 Kobe 0.2 El Centro 0.2 1.255 2.255 06070309 step impulse 2.2 1. North Date Code Type of Test opening opening width [m] width [m] 06062304 step impulse 1.2 06062811 Nocera Umbra 0.2 1.255 2.255 2.15g 1.2 06062311 white Noise 1.2 1.15g 1.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.2 06062806 1. Replacement of screws in vertical joints of walls 1S 1N 2S 2N.2 1.255 06070306 El Centro 0.255 06070308 Kobe 0.2 06062808 white Noise 1.2 1. retightening of holdown bolts. 06062809 step impulse 1.255 2.255 2.2 1.255 2.50g 06062807 step impulse 1.50g 06062812 step impulse 1.2 1.2 06062309 Nocera Umbra 0.60g 06062313 step impulse 1.2 1.15g 2.15g (interrupted) 1. Portugal November.2 1.255 06070311 2.2 1.2 06062305 white Noise 1.2 st Enlargement of openings in South and North wall at 1 Floor.2 06062314 white Noise 1.255 2.255 2.2 1.2 06062810 white Noise 1. 1st F.50g 06070312 step impulse 2.2 1.2 Kobe 0.2 1.255 2. 06070303 step impulse 2.2 06062306 Kobe 0.2 Replacement of screws in vertical joints of walls 1S 1N 2S 2N.255 06070310 white Noise 2. tightening of holdown bolts.2 06062307 Kobe 0.2 1.2 1.2 1.2 06062310 step impulse 1.2 06062312 1. South 1st F.15g 2.255 2.255 Configuration A Configuration B TESTS 2006-07-03 TESTS 2006-06-28 TESTS 2006-06-23 1009 89 .255 06070307 Nocera Umbra 0. Movement of 4 relative displacement transducer.2 06062813 white Noise 1.255 Kobe 0.255 06070313 white Noise 2.15g 2.2 06062805 White Noise 1. 2006 Table 4 Steps of shaking table tests.255 06070304 white Noise 2. 06062804 step impulse 1.15g 1.2 1.2 1. retightening of holdown bolts.2 Replacement of screws in vertical joints of walls 1S 1N 2S 2N.2 1.2 06062308 El Centro 0.255 2.15g 2.
North opening opening width [m] width [m] 2.5g 06070511 step impulse 4. Movement of 2 relative displacement transducer.255 2.5g 06070306 step impulse 2. 06070513 step impulse 4.255 Configuration C Replacement of screws in vertical joints of 1st and 2nd storey walls.255 2.15g 4.255 2.255 Replacement of screws in vertical joints of walls 2S 2N.15g 4. 06070301 step impulse 2.00 2.255 2.00 2.255 2.255 06070509 white Noise 4.00 2.255 06070311 Nocera Umbra 0.255 06070508 step impulse 4.00 2.255 El Centro 0.255 06070512 white Noise 4.255 06070506 El Centro 0. 06070309 step impulse 2.255 2.00 2.00 2. TESTS 2006-07-05 TESTS 2006-07-04 90 .255 2. Adding of 2 new holdown anchors at 2nd floor in positions corresponding to the top corner of the 4m opening.00 2.00 2.15g 4.255 2.255 El Centro 0.00 2.255 06070517 white Noise 4. retightening of base holdown bolts. Replacement of screws in vertical joints of walls 1S 1N 2S 2N.30g step impulse white Noise 1st F.5g 06070516 step impulse 4.255 Retightening of holdown bolts.00 2. retightening of holdown anchor bolts. Portugal November.255 4. retightening of holdown bolts.255 06070310 white Noise 2. 2006 Date Code 06070314 06070315 06070316 Type of Test El Centro 0.255 06070504 4.255 2.255 06070307 white Noise 2.255 06070515 4.00 white Noise 2.255 st Enlargement of openings in South wall at 1 Floor.255 06070313 white Noise 2.00 2.255 06070302 white Noise 2. South 1st F.255 06070514 white Noise 4.255 2.255 2.255 06070505 Kobe 0. 06070503 step impulse 4.50g 06070312 step impulse 2. retightening of holdown bolts.00 Nocera Umbra 0.255 06070305 2.255 Replacement of screws in vertical joints of walls 2S 2N.00 2.255 2.255 2.255 06070510 2.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Adding of 4 screws for each vertical joint over the 4m opening.255 06070507 Nocera Umbra 0.00 2.255 2.255 2.255 2.
2g 06071010 step impulse 4. North opening opening Date Code Type of Test width [m] width [m] 06070704 step impulse 4.00 2. 06070709 4.00 2.255 06070708 white Noise 4.00 Kobe 0.00 2. 06070712 step impulse 4.255 06071005 white Noise 4.255 06070711 white Noise 4.00 2.5g (low resolution) 06070720 step impulse 4.00 2.00 2.00 2.8g 06070710 step impulse 4.00 white Noise 2.open test Retightening of holdown bolts.00 2.255 06071008 white Noise 4.255 Replacement of screws in vertical joints of walls 2S 2N.00 2.255 Kobe 0.00 2. 06070717 step impulse 4.80g .00 2.255 Kobe 0.00 2.50g (interrupted) 06070707 step impulse 4.255 Nocera Umbra 1.255 TESTS 2006-07-10 TESTS 2006-07-07 1011 91 .255 nd Retightening of holdown bolts at 2 floor and of screws in vertical joints of walls at 1st and 2nd storey.255 Kobe 0.255 06071011 white Noise 4.00 2.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.255 06071006 4.00 2.255 Kobe 0.255 06070719 4.00 2.255 06070706 2. 06071004 step impulse 4.255 06070716 white Noise 4.00 2.00 2.255 06070718 white Noise 4.255 06070713 white Noise 4. Portugal November.255 4.00 2.255 Replacement of screws in vertical joints of walls 2S 2N.255 06071009 4.00 2. retightening of holdown bolts.255 06070721 white Noise 4.00 2.255 06070714 4. Replacement of screws in vertical joints of walls 2S 2N. 06071007 step impulse 4.00 2.50g (high resolution) 06070715 step impulse 4. retightening of holdown bolts. South 1st F.00 2. 2006 1st F.00 2.00 2.255 Retightening of holdown bolts.255 06070705 4.
50 0.56 35. the torsion of the building was counteracted by the contribution of the two perpendicular walls.79 4.07 37. Comparison of frequencies between Phase A and Phase C shows that even with larger openings in Phase C (a total of 8.88 5.57 5.19 11.02 8.50 0.50 0.05 5. In table 6 an overview of the observed damage is shown only for Configuration C.76 5.18 4. 92 . In table 5 the test results are summarized for the ground motion records of 0.20 7.30 4.50 0.59 4.96 5.50 0.27 5.88 4.00 29. Test results The tests results are summarized in table 5 and 6 only for the most important quakes.51 9.80 1.90 13.15 9. 2006 3.79 4.60 0.90 23.57 5.63 12.20 4. Table 5 Summary of results in terms of 1st storey drift and natural frequencies.30 4. any damage that couldn't allow for any further reparation of the building. Of course with the most severe earthquakes (those listed form the thick black line on) a major damage was observed but nevertheless the building was still repairable.37 5.97 19.50 0.00 As could be noted the asymmetric configuration of the building in Configuration C didn't lead to almost any pronounced torsional behaviour.30 4.50 0.47 5.88 4.79 4.30 4. Configuration Code Record PGA (g) First storey drift [mm] North South A A A B B B C C C C C C C C 06062312 Kobe 06062806 El Centro 06062811 Nocera Umbra 06070311 Kobe 06070305 El Centro 06070311 Nocera Umbra 06070510 Nocera Umbra 06070515 El Centro 06070706 Kobe 06070709 Kobe 06070714 Kobe 06070719 Kobe 06071006 Kobe 06071009 Nocera Umbra 0.65m of Phase A) the specimen didn't show a significant reduction.08 Measured Natural Frequency [Hz] Before 6.08 4. the specimen didn't show a significant damage (5% in Phase A.68 34.67 7.96 20.86 5.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.5m of openings in the walls parallel to the shaking direction compared to the 4. i. Portugal November.02 25. which was the configuration in which a near-collapse status was reached.e.80 0.50 0.66 5. 6% in Phase B and C).36 13.56 6.50 21.50 0.71 13. An accurate analysis of the trend of frequencies shows also that in all the three phases between the beginning and the end of the series of the three 0. As a first comment it should be said that the test house survived 15 destructive earthquakes without any severe damage. This may be explained by the fact that as the floors were rigid enough.95 12.01 10.50 0.07 4.54 10.55 10.5g or more for each configuration in terms of first storey drift for the walls and of measured natural frequency before and after each test.90 11.10 38.5g earthquakes that were scheduled initially.30 After 5.39 8.07 5.35 6.
Portugal November. 2006 Table 6 Results of shaking table tests for Configuration C in terms of observed damage. connected to the basement by means of 8. Kobe 06070714 0. SIMPSON STRONG-TIE holdown anchors HTT22.1 Design the structure using q=1 according to the seismic code for a given design PGAu.code.50 06070719 0.50 Tightening of holdown anchor bolts.80 Idem Kobe Kobe Kobe Kobe 06070714 0. Analyze the test results and calculate q as the ratio between the PGAu.eff value that caused the "actual" near-collapse of the building and the design value of the PGAu.code (0.8 Class M16 anchor bolts and with 12 ø4 annular ringed nails to the cross-laminated walls.1.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.35g which is the design ground acceleration corresponding to the more hazardous seismic zone of Italy) Define.1 Design of holdown anchors at ground level To resist the shear forces steel angles have been used and calculated. as above mentioned. The distribution of holdown anchors at the ground floor and of the seismic forces at each floor is showed in Fig. as "near-collapse" criterion the failure in holdown anchors (one or more). To resist the uplifting forces holdowns anchors have been used and calculated. Replacing of screws in vertical joints between panel Tightening of holdown anchor bolts. as before mentioned.50 06071006 0. Replacing of screws in vertical joints between panel Record Code Observed damage (after the test) None None None Slight deformation of screws in vertical joints between panels None None Slight deformation of screws in vertical joints between panels Holdown failure and deformation of screws in vertical joints between panels Nocera Umbra 06071009 1.20 The definition of a collapse criterion (the failure of one ore more holdown anchors) could allow the calculation of the q-value at least for one earthquake using the following strategy: • • • 3. 11.80 Idem Tightening of holdown anchor bolts Replacing of holdown anchors and tightening of bolts. The holdown anchors used to connect the building at ground floor were. 1013 93 .50 Tightening of holdown anchor bolts Replacing of screws in vertical joints between panel 06070709 0. Evaluation of the q-value for Nocera Umbra earthquake 3. PGA Restoring intervention (before the test) [g] Nocera 06070510 0.50 Tightening of holdown anchor bolts Umbra Replacing of screws in vertical joints between panel El Centro 06070515 0.
according to Eurocode 5 and 8. the strength design value of each nail is: Rd = Rk × k mod γM = 4 × 1.93 − 5 × H × 6. Portugal November.40m Considering only the design of the holdown anchors at the ground floor and taking into account also the contribution of the holdowns in the walls perpendicular to the shaking direction the calculation gives the following results (moment equilibrium around the A line and neglecting the contribution of holdown at openings): Fr × hr + F2 × h2 + F1 × h1 − W × 6. 2nd floor and roof are respectively 3.34 kN (6) Note that the design tensile strength of the Class 8.8 φ16 anchoring bolt.09.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.1 = 3.38 kN 1.40 + 279 × 6. considering the effective cross section is: 94 . which is taken as the 5-percentile value of strength. Therefore.93 = 0 2 6. 2006 Fig.93 = 0 Ö 2 H =40.09 − 465 × From the results of the experimental tests on the steel to timber connections using annular ringed nails.18 and 9. Following the procedure described in 2.3 (5) Hence to resist the uplift force each holdown anchor shall be connected using 12 nails.2 the calculated seismic forces are the following: Fr = 91 kN F2 = 279 kN F1 = 139 kN The heights of the 1st floor. 11 Distribution of holdown anchors and steel angles at ground floor and distribution of seismic forces at each floor. 6.93 − 5 × H × 6.56 kN > H = 40. each nail has an ultimate shear resistance of 4 kN. Hr = 12x3.18 + 139 × 3.34 kN (3) (4) 91 × 9. In the left side picture the holdown anchors marked with H are those taken into account in the design.38 = 40.
1. [1] [2] [3] References EN 1998-1:2004: "Eurocode 8: Design of structures for earthquake resistance . M.P.20 = 3. Acknowledgments A very special thank for their fundamental contribution to the success of this research to Prof.code equal to 0. Firenze 2006 H. 5.Part 1: General rules.2. Japan. M. X.2g test.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Follesa. 12 Holdown failure after Nocera Umbra 1.39-15-4. This is obviously impossible. H. Lu. M.35 kN which is greater than Hr 1000 × 1. Kawai. the q value is: q= 1.2 Evaluation of the q-value Being the design ground acceleration PGAu. to Gabriele Bonamini. Discussion and conclusion Of course the above value is valid only referring to the used Nocera Umbra ground motion record. Proceedings of 39th CIB W18 Meeting. seismic actions and rules for buildings" A.35g. Motoi Yasumura of Shizuoka University. Lauriola. Karacabeyli "Research program on the seismic resistance of conventional wood-frame construction". Sandhaas.35 (8) 4.4 0. Mario Moschi. Japan. by applying the procedure given in 2. J. San Francisco 2006 1015 95 . C. Cheng H. Fig. paper n. Ni. Yasumura: "Which Seismic Behaviour Factor for Multi-Storey Buildings made of Cross-Laminated Wooden Panels?". N. Portugal November. C. Mario Pinna and Carmen Sandhaas of CNR-IVALSA and to Masateru Soudou of the Tsukuba Building Test Laboratory of Center for Better Living. M. to Dr Chikahiro Minowa of NIED and to Marco Pio Lauriola of CNR-IVALSA and Minoru Okabe of the Tsukuba Building Test Laboratory of Center for Better Living. Japan without which this test program wouldn't have been possible. A series of different quakes should be used with the same procedure. Ceccotti. Proceedings of 8th National Conference on Earthquake Engineering (NCEE). Rainer. Naohito Kawai of BRI. therefore the importance of a good mathematical model that can simulate different quakes and cases. Minowa.1 (7) 3. 6. E. Follesa. C. 2006 Nr = Ares × f y γM = 157 × 640 = 91. Thanks also for their contribution to Dr.
Teaching Assistant for Building Materials and Experimental Analysis of Structures.1). Introduction Wooden panel structures have a general reputation to perform well when subjected to strong earthquakes because of: large strength-to-weight ratio of wood. from 2001. Ph.. The new trends are bringing also multi-storey structures. but there are also cases where the behaviour of such buildings was not as expected (Fig.D. Experimentally based modelling of structures. Eng. high degree of structural redundancy and energy dissipation. from 1992. However. Born 1950. Therefore. Civ. Therefore herein the importance of experimental investigation of earthquake response of anchored massive wooden panels for earthquake safe design of wooden buildings is addressed. 97 . timber construction itself is not a guarantee for adequate seismic performance. the appropriate guidelines for designing have to be set for existing and new timber structural systems to assure their seismic resistance. analytical and experimental research in materials and seismic resistance of timber and composite structures. The reputation of good behaviour had been proved in many lessons learned from post earthquake observations. The dynamic tests have proved the ductile behaviour of the connections at both models and exhibit good correlation with the results from quasi-static tests. Civ.si Roko ZARNIC Ph.D.si Born 1971.uni-lj. Description of cyclic testing of single panels and testing of panel assemblages on shaking table is presented to explain the benefits of experimentally obtained data for the design practice. It is obvious that there is a need for development of an integral European standard that would cover cyclic testing of construction parts and wall segments.D. Ph. Summary The new generations of massive cross-laminated ("X-lam") wooden structures are recently becoming more popular in European market. 2006 STUDY OF LATERAL RESISTANCE OF MASSIVE X-LAM WOODEN WALL SYSTEM SUBJECTED TO HORIZONTAL LOADS Bruno DUJIC Ph.D. Teaching Assistant University of Ljubljana Faculty of Civil and Geodetic Engineering Ljubljana. Associate Professor University of Ljubljana Faculty of Civil and Geodetic Engineering Ljubljana. Chair of Research in Materials and Structures..uni-lj. Professor of Building Materials and Experimental Analysis of Structures. Portugal November. Slovenia bdujic@fgg. Eng. 1. The new standard should also include the criteria for determination of limitations of inter-story drifts according to the concept of performance based earthquake engineering design. The development of experimentally supported knowledge about the response of solid wooden houses is illustrated by description of test procedures and brief information about results of testing. Special attention is paid to buildings located in earthquake prone areas of middle and south Europe. The importance of a proper taking into account of the boundary conditions and of the influence of vertical and type of horizontal loading is evident from comparison of the behaviour of differently tested panels. Slovenia rzarnic@fgg. expertise in repair and strengthening of structures in earthquake prone areas.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Experiences of the recent earthquakes show that damages and even collapses of wooden buildings do not endanger the human lives in the same extend as it is the case with the buildings constructed from massive materials as concrete or masonry (Table 1). enhanced strength under short term loading.
The post earthquake observations of damaged wooden houses and analysis of experimentally tested structural elements developed the world wide knowledge about response of wooden buildings on earthquake and strong wind. Properly designed building encompasses the benefits of all these parameters and their interrelations. 2. timber rivets or nails should be used for connection of wooden panels to enable ductile behaviour of the entire building.2: Seismic activity in Europe 1973 2002. connections should be weaker than the members they connect. Slender bolts.1 6.000 50. the ability of structural elements and whole building to dissipate energy imposed by earthquake diminishes the destructive action of earthquake.300 Woodframe <10 4 0 0 0 16 + 4 0 Alaska.4 6. Learning from the on-site observations and experimental investigation researchers have developed different test protocols and test setups trying to simulate the natural behaviour of buildings as realistic as possible. Element strength makes building resistant to seismic forces and prevents development of damages up to certain limit. global failure because of the effect of soft storey.7 7. In the case of wooden buildings appropriate design of joints and anchorages gives a structure high potential in dissipation of energy and thus increased resistance to earthquake action. In addition.8s.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. All these parameters are mutually inter-related. dowels. Magnitude > 3 [2].2s to 0. strength and ductility both on the level of the constituent structural and non-structural elements and whole building.3 5. Seismic resistance of wooden building The seismic resistance of buildings is governed by three main parameters: stiffness. ZDA. Typical wooden structures have fundamental period T of 0.000 10.7 6.000 8. A significant part of the European territory is endangered by earthquakes (Fig. but the interaction of structural and non-structural elements of different ductility should be carefully studied.8 Fig. 1964 San Fernando. observing the behaviour of wooden structures one have to bear in mind that the wood is an anisotropic material where strength and stiffness properties vary with the orientation of the wood fibres. Ductility enables reduction of seismic forces. but attracts the increase of forces that are proportional to stiffness.2) where the contemporary wooden houses are becoming widely popular because of their attractiveness for residential and business use. Earthquake Magnitude Buildings Shaken Persons killed (Approximate) Total 130 100.000 63 0 0 66 60 6. while in the case of platform frame are in range between 3 to 20%. Typical damping values of heavy timber are in the range between 3 and 5%. However.1: Northridge. Portugal November. 1988 Loma Prieta. 1971 Edgecumbe. 1987 Saguenay. The most important parts regarding the earthquake response of timber structures are connections and anchors of wooden panels. Resonance likely occurs on softer soils. 1994.7 6. Hyogo-ken Nambu. 1995 Fig. 2006 Table 1: Casualties in recent earthquakes. One of the major problems of understanding is related to boundary conditions and influence of vertical loading on building elements.000 7. Stiffness of structural element provides limitation of the lateral deflections.000 200. 1994 8. According to the basic principles of capacity design. In the case of wooden structures the connections that stress the wood perpendicular to grain should 98 . During an earthquake stiffness of structure is decreasing due to softening of joints and anchors what increase the fundamental periods. 1989 Northridge.
With establishing such relationships dynamic properties also for other type of light load-bearing systems could be defined on the base of known quasi-static cyclic or monotonic mechanical properties. Dynamically developed failure mechanisms on wall segments could confirm reality of proper boundary conditions set at quasi-static racking tests executions.4). In the scope of the project numerous of racking tests were carried out at University of Ljubljana. Differences in mechanical properties between monotonic and cyclic responses were studied. The fastener shank is the only source Deformation of connection resistance. The loops are mildly to severely pinched (Fig. Wall panels were tested at various boundary conditions which enabled wall deformations from cantilever up to pure shear. Load The hystertic lines are usually smooth Virgin curve withouth well defined yield point. Connector moving throught gap Pinching effect is caused by the loss of (reduced stiffness) stiffness at small deformations. Experimental research of X-lam solid wooden walls panels The research reported in this paper has been carried out within the project supported by Austrian Company KLH Massiveholz GmbH. 3.3). Portugal November. Faculty of Civil and Geodetic Engineering on walls of length 244 cm and story height at combined constant vertical load and monotonous or cyclic horizontal load applied according to different loading protocols. Hysteretic behaviour of connections and anchors are governed by the material properties of wood and fasteners. Macedonia in order to investigate the response of massive X-lam wooden wall system under earthquake excitation. The behaviour of building during an earthquake can be predicted by an Connector mathematical models that usualy depends on regaining contact various parameters that can only be defined by (wood crushing) Strength experimental investigation of constituent deterioration elements of wooden building starting from cyclic testing of connections. It can be achieved by simulation of realistic boundary conditions and taking into account the influence of magnitude of vertical load and influence of horizontal loading protocol In this paper the brief insight in cycling and shaking table testing of anchored solid. by shaking testing of structural assemblages and/or prototypes of entire buildings. magnitudes of vertical load and type of anchoring systems were evaluated on the base of deformation mechanisms and racking strengths of wall segments. Additionally two full-scale one story models have been constructed and tested on the shaking table at the IZIIS Laboratory in Skopje. Testing protocols may be standardized or developed by researchers. cross-laminated wooden panels is presented. Influences of boundary conditions. The main target of the project was to obtain reliable data on the mechanism of behaviour of massive X-lam wooden wall panels that will contribute to design and construction of seismic resistant and safe buildings. Comparative tests on structural segments with two parallel walls of length 244 cm and story height linked together with wooden ceiling were dynamically tested with harmonic and seismic excitation.1 Cycling testing of panels Following the experiences obtained from testing of masonry panels. followed by Fig. 3. The main task of testing is simulation of realistic response mechanism. 1003 99 . A cavity around the fastener is formed by wood crushing.3: Hysteretic behaviour of ductile connection cyclic testing of anchored panels and ending capable to disipate energy. a universal shear wall test setup was developed and installed at UL FGG in 1999 (Fig. 2006 be avoided in order to mitigate damages due to splitting of wooden elements.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. The main task was to define mechanical properties of dynamic responses of tested models. The basic idea was to make a correlation between the results from the quasi-static tests and the results from the shaking table tests.
International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. The main idea of the device is to use a gravity load induced by ballast as a constant vertical load and a displacement controlled hydraulic actuator as a driver of the cyclic horizontal load. 2006 Fig. where one edge of the panel is supported by the firm base while the other can translate and rotate as much as allowed by the ballast that can translate only vertically without rotation ("Case B") shear wall mechanism.5: Typical responses of wooden wall panels exposed to combined vertical and horizontal load. what is an insufficient tool for 100 . where one edge of the panel is supported by the firm base while the other can freely translate and rotate ("Case A") restricted rocking mechanism. The complete information about the mechanical characteristics of wooden wall panels and their anchoring can be obtained from responses both to monotonous and cyclic loading with proper combination of vertical forces (Fig. The "Case B" is typical for panels carrying the floor construction above it and the "Case C" is the typical case of infill embedded in the stiff surrounding frame. Following this idea. In "Case A" and "Case B" the panel is exposed to constant vertical load in every stage of the cycling excitation or horizontal deformation induced along the upper edge where the ballast is acting. The main challenge is to simulate realistic boundary conditions that may occur during the action of an earthquake. the testing device should allow the altering of boundary conditions from one to another test run. In reality. Basically. In practice. the "Case A" represents mostly the behaviour of narrow panels and panels located in attic and vertically loaded only by flexible roof constructions. Portugal November.4: Setup for cycling testing of panels. the boundary conditions may change during an earthquake excitation because of changes of the building characteristics due to development of damages.6). In "Case C" the vertical load increases when the panel wants to uplift due to displacements along the upper horizontal edge. the protocol of EN 12512 covers only cyclic testing of particular joints made with mechanical fasteners. where one edge of the panel is supported by the firm base while the other can translate only in parallel with the lower edge and rotation is fully constrained ("Case C") Combined shear . Therefore. The protocol of EN 594 is sufficiently covering the monotonous loading of wall panels. three major cases of boundary conditions are most likely to appear in reality (Fig. the set-up can be easily adapted to various boundary conditions applied to tested panels. Unfortunately.rocking Rocking response of walls Shear response of walls response of wall Fig.5): shear cantilever mechanism. The advantage of the testing procedures proposed by us following the "Case A" or "Case B" is avoiding the boundary conditions of the "Case C".
44 m. However. The cyclic response shows higher initial stiffness due to hardening of the fasteners exposed to low-cycle fatigue and lower ductility (down to 50% of the ductility reached by monotonous loading).Case C 20 10 0 -60 -50 -40 -30 -20 -10 -10 -20 cyclic horizontal load -30 -40 -50 EN 594 defined horizontal load monotonous incrising horizontal load 0 10 20 30 40 50 60 90 60 Boundary Conditions Case A Fmax (Case C) Fy (Case C) Fmax (Case A) Fy (Case A) 30 0 10 15 20 25 30 35 Vertical load per meter length of the wall [kN/m'] Displacement at the top of the wall [mm] Fig. 1005 101 .5 m in both orthogonal directions. cyclic testing does not give us all information about behaviour of structural elements during an earthquake. The reason therefore is that ultimate joint displacement is used.6: Comparison of monotonic and cyclic response of solid wooden wall panels (length of 2.7: Influence of vertical load intensity on load carrying capacity of solid wooden wall panels with length of 2. The necessary information about earthquake response of anchored panels can be completed only by the shaking table testing of full scale panels as described in continuation of this paper. support the table. Skopje.8). In the case presented in the diagram the load carrying capacity of the panel exposed to cyclic loading was 10 to 20% lower than the resistance of the panel exposed to monotonous loading. Comparison of the responses of wooden panels subjected to cyclic and monotonous loading (Fig. 2006 evaluation of the behaviour factor "q" needed for design of earthquake resistant buildings. On the other hand.2 Shaking table testing of anchored panels Two full-scale models have been constructed and tested on the shaking table at the IZIIS Laboratory.44 m) vertically loaded with 15kN/m'. 50 40 150 Hor. Four vertical hydraulic actuators located at four corners. instead of yield slip (EN 12512) which is difficult to define. The new standard should also include the criteria for determination of limitations of inter-story drifts according to the concept of performance based earthquake engineering design. at a distance of 3. In the case of the stiff solid wooden panel behaviour shear mechanism does not develop in spite of varying boundary conditions from the "Case A" to the "Case B".6) illustrates the importance of cycling testing. It is obvious that there is a need for development of an integral European standard that would cover both monotonous and cyclic testing of wall diaphragms. Macedonia in order to investigate the response of massive wooden wall panel systems under earthquake excitation. Therefore.0x5. The total weight of the shaking table is 330 kN. the ISO 16670 standard also addresses only the joints but the proposed protocol can be also used for testing of wooden wall diaphragms. The graphs in Figure 7 clearly illustrate the influence of vertical load intensity both on the load carrying capacity and the type of response mechanisms as discussed above. load carring capacity [kN] 30 120 Horizontal Load [kN] Boundary Conditions .International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.0 m in plan (Fig. The shear mechanism was reached only when the boundary conditions were set to the "Case C". 3. earthquake design of wooden buildings can not be properly performed without data obtained from cyclic testing of panels exposed to different intensities of vertical load. Fig. Portugal November. The shaking table on which the structural models were installed in order to be subjected to a biaxial earthquake motion is a pre-stressed reinforced concrete plate 5. Since the ISO protocol is based on ultimate displacement it can forward a behaviour factor "q" as addressed in Eurocode 8. with total force capacity of 888 kN. The basic idea was to make a correlation between the results from the cyclic tests preformed at the UL FGG in Ljubljana and the results from the shaking table tests.
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
The maximum applied accelerations are: vertical 0.50 g and horizontal 0.70 g with maximum displacement in vertical direction ±0.050 m and in horizontal direction ±0.125 m. The frequency range is 0-80 Hz. The total bearing capacity for static loads is 720 kN. The shaking system controls five degrees of freedom of the table, two translations and three rotations. The analog control system controls displacements, velocity, differential pressure and acceleration of the six actuators. A complete package of computer programs for control and Fig.8: Shaking table installed at IZIIS in Skopje, Macedonia. acquisition has also been used. Both tested specimens were made of cross-laminated wooden panels, produced by the Austrian Company KLH Massiveholz GmbH. Specimen 1 was assembled by two one-unit wall elements 244/272/9.4cm and Specimen 2 by two wall elements of half width screwed together in wall panel. Both of the models had a roof element of size 244/210/16.2 cm. Models were stiffened by two additional panels 190.5/272/9.4 cm placed in lateral direction (Fig.9). Tested models were instrumented by LVDTs and accelerometers. LVDTs were used for measuring of horizontal displacements of the top of models, slipping between the wall and roof element, diagonal deformations, uplift of wall and slip between the wall panel and RC foundation. In addition, Specimen 2 was instrumented also for measuring of the slip between two screwed single panels. Accelerometers were used for measuring of accelerations of shaking table, RC model foundations, top of models, and atop of each panel to check the torsional movements.
Fig.9: Instrumented specimens fixed on shaking table and ready for testing.
Fig.10: Detail of wood panel foundation anchor.
Wooden panels were anchored in RC foundation beams by ribbed angle irons of 105 mm that were on every 60 cm of wall length. Irons were fixed KLH panels by 10 annularly nails 4,0/60 mm and to the RC foundation beams by two bolts M12 (Fig.10). The ballast applied atop the each model corresponded to the 3-storey structures. Preliminary estimations defined a mass of 5 tons acting on single wall panel with length of 244 cm representing ground floor wall of the considered building type. Following this estimation and taking into consideration that each model consisted of two wall panels and one roof panel additional mass of 9.6 tons was applied. 24 steel ingots (3 layers by 8 ingots, 400 kg each) were placed and connected rigidly to the roof panel (Fig.9). Test procedure was equal for both models after their installation on shaking table and loading with ballast. It started with measuring of the fundamental period of models by the ambient and low level
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
random vibration tests. The results of both test matched well. Natural frequency of 3.4 Hz in lateral and 7.4 Hz in longitudinal directions for Specimen 1 and 3.6 Hz and 7.4 Hz respectively for Specimen 2 were obtained by ambient vibration test. Several test runs were carried out during shaking table test. The peak ground acceleration increased gradually from 0.06g to 0.38g. Following earthquake excitation has been applied: El Centro 1940, N-S, California, USA, Petrovac 1979 Montenegro, Kobe 1995 E-W, Japan, and Friuli 1976 E-W, recorded in Tolmezzo, Italy. All earthquakes were applied with real intensity only Kobe earthquake was limited to scaled factor 0.5 due to technical reasons. After excitations by earthquakes models were also subjected to harmonic excitation with frequencies of 7.5Hz and 5.0Hz. Each earthquake and harmonic excitation was followed by the low-level random vibration tests in order to monitor the change of the natural frequencies due to potential eventual damages of specimens. Maximal acceleration responses have been obtained from Petrovac excitation. The displacements of the shaking table were up to 68.75mm, and maximum obtained accelerations of models were amax=0.6g for Specimen 1 and amax=0.5g for Specimen 2. Due to the increased energy dissipation between the screwed panel elements Specimen 2 demonstrated far more ductile behaviour than Specimen 1. On diagram in Figure 12 relative displacements measured at the top edge of the walls were compared as responses of Specimen 1 and and Specimen 2 on El Centro earthquake record. Relative displacement of Specimen 2 was 42% higher than relative displacement of 6 S pec im en 1 Specimen 1 while acceleration at the top 4 S pec im en 2 was 57% higher. Relative displacement in 2 presented response reached 0.4% of the story height or H/250, where H is wall 0 height. In future, the recorded data and 0 5 10 15 20 25 30 -2 comparison of test responses will yield in -4 set of parameters describing the influence of connection details. -6 -8 In general, shaking tests proved the nonlinear behaviour of the massive wooden Tim e (s) wall panel systems. The main source of the Fig.11: Comparison of the measured acceleration non-linearity is certainly the connection, response of both models excited by Petrovac 1979, i.e. the steel anchorage system (Fig.10). Montenegro earthquake. The solid wooden wall panel itself behaves 10 linear-elastically for most of the cases and, 8 of course, as an orthotropic material. 6 Connections are much weaker part of the 4 system and thus enable high level of 2 energy dissipation. Consequently, the 0 -2 entire non-linear behaviour results from -4 the behaviour of nailed connection of -6 anchor angle iron (Fig.10) to wooden Specimen 1 -8 panel. The vertical bolts anchored into the Specimen 2 -10 concrete foundation together with the steel -12 0 5 10 15 20 25 30 35 40 45 plate form a very stiff part of the Time [s] connection, so that we cannot expect that Fig.12: Comparison of the relative horizontal displ. these components contribute to any nonof both models measured at the top edge of the walls linear vertical and/or horizontal relative during the El Centro 1940 earthquake excitation displacements between the RC foundation where shaking table displ. were up to 106.25 mm. and the wooden wall.
Relative Displacement [mm]
Acceleration (m /s**
The development of experimentally supported knowledge about the response of solid wooden houses is illustrated by description of test procedures and brief information about results of testing. The research program starts with testing of elements and joint details, follows with cycling testing of anchored wooden walls and concluding with shaking table resting of wall assemblages. The importance of a proper taking into account of the boundary conditions and of the influence of
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
vertical and type of horizontal loading is evident from comparison of the behaviour of differently tested panels. There is a need for further development of standard protocols for wooden wall diaphragms used for structures located in earthquake prone areas. New standards should implement the concept of performance based earthquake engineering design to obtain experimental data needed for evaluation of the behavior factor "q" and to set the values of story drifts defining the limit states of the story base shear diagram. Paper presents the first brief information about shaking table tests of simple wooden panel assembly where anchorage and inter-panel connection systems were of prime interest of testing. Two full scale wooden wall assemblies were tested named as Specimen 1 and 2. Both were made of cross-laminated wooden panels, produced by the Austrian Company KLH Massiveholz GmbH. Specimen 1 was assembled by two one-unit wall elements 244/272/9.4cm and Specimen 2 by two wall elements of half width screwed together in wall panel. Both of the models had a roof element of size 244/210/16.2 cm. Models were stiffened by two additional panels 190.5/272/9.4 cm placed in lateral direction Series of dynamic tests have been preformed in order to investigate the structural behaviour of massive wooden wall system under dynamic conditions. Several earthquake records, significant for Central and Southern Europe have been applied to the models. Maximum acceleration response have been recorded for Petrovac earthquake excitation, amax=0.6g for Specimen 1 and amax=0.5g for Specimen 2. Both of the test specimens behaved according to expectations and no visible damage was registered. In both models seismic energy has been dissipated through panel to foundation anchorage system. In addition, the second specimen dissipated energy also through vertical screwed connection between two wall units. The dynamic tests have proved the ductile behaviour of the connections at both models and exhibit good correlation with the results from quasi-static tests.
The research carried out in UL FGG Laboratory was industrial project supported by Austrian Company KLH Massiveholz GmbH and a part of Slovenian national research project on improvement of seismic resistance of wooden precast buildings. The Ministry of Science, Higher Education and Technology of Republic of Slovenia financed this three year project in period 20032005. The research cooperation with IZIIS was enabled by the bilateral Slovenian and Macedonian cooperation supported by Ministry of Science, Higher Education and Technology of Republic of Slovenia. The support of all projects' financers is gratefully acknowledged.
[1] [2] [3] [4] [5] [6] [7] [8]
Rainer, J.H. and Karacabeyli, E. (1999). Wood-Frame Building Construction in Earthquakes. Forintek. Canada Corp., Special Publication SP-40, Vancouver ENV 1995-1-1:1993. Eurocode 5 – Design of timber structures – Part 1-1: General rules and rules for buildings, CEN, European Committee for Standardization, Brussels. EN 594:1995, Timber structures - Test methods - Racking strength and stiffness of timber frame wall panels, CEN, European Committee for Standardization, Brussels. Dujic, B., (2001). Experimental Supported Modelling of Response of the Timber-Framed Wall Panels to Horizontal Cyclic Load. Ph.D. Thesis (in Slovenian), UL FGG, Ljubljana, Slovenia. Dujic, B., and Zarnic, R. (2002). Influence of Vertical Load on Lateral Resistance of TimberFramed Walls. Proceedings of CIB-W18/35-15-4, Kyoto, Japan. Dujic, B., Pucelj, J., Zarnic, R., (2004) Testing of Racking Behavior of Massive Wooden Wall Panels, Proceedings of CIB-W18/37-15-2, Edinburgh, Scotland. Dujic, B., Aicher, S., Zarnic, R., (2005). Investigation on In-plane Loaded Wooden Elements – Influence of Loading and Boundary Conditions. Otto Graf Journal, Materialprüfungsanstalt Universität Stuttgart, Otto-Graf-Institut(FMPA), MPA Stuttgart; Vol.16 2005. Hristovski, V., Stojmanovska, M., (2005). Experimental and Analytical Evaluation of the Racking Strength of Massive Wooden Wall Panels -Preliminary Project Phase, EE21C Skopje-Ohrid, Macedonia.
International Workshop on "Earthquake Engineering on Timber Structures"
Coimbra, Portugal November, 2006
Bruno DUJIC Ph.D., Civ. Eng. Teaching Assistant University of Ljubljana Faculty of Civil and Geodetic Engineering Ljubljana, Slovenia firstname.lastname@example.com Simona KLOBČAR B.Sc., Civ. Eng. Postgraduate Student Company TRIMO d.d. Trebnje, Slovenia email@example.com Roko ŽARNIĆ Ph.D., Civ. Eng. Associate Professor University of Ljubljana Faculty of Civil and Geodetic Engineering Ljubljana, Slovenia firstname.lastname@example.com
In structural design of wooden structures the load-bearing capacity of fenestrated wall segments usually is not taken into account when earthquake resistance analysis is carried out. Some experimental and parametric studies show that fenestrated wall segments may significantly contribute to earthquake resistance of wooden buildings. The experimental research was mainly done on timber frame walls with plywood sheathing (Sugiyama, Yasumura, Dolan, and Johnson). The load-bearing capacity and stiffness of fenestrated wooden walls are influenced by dimensions and layouts of openings. The aim of the experimental research and parametric study presented here is to provide the first information about estimation of racking strength and stiffness of crosslaminated solid wooden walls with openings and to recognize how shape and area of an opening influence on reduction of shear properties compared to full cross-laminated wooden wall. Preformed parametric study resulted in diagrams that could serve for engineering judgment of fenestration influence on stiffness and load-bearing capacity of cross-laminated solid wooden wall. It was concluded that regularly positioned openings in surface equal up to 30 % of wall surface do not much influence on load-bearing capacity when stiffness is reduced for about 50 %.
Shear walls are the main structural elements used to resist seismic and wind loads in timber structures. Currently, timber buildings are designed for earthquake and wind loads by taking into account in structure shear resistance only full wall segments. Parts of wall where different openings appear are excluded from calculation method (Fig.1). This simple design procedure usually causes underestimation of shear resistance of a wooden building. For more accurate and economic design parts of wall above and below the openings have to be taken into account as construction parts which transfer loads between full wall segments and dictate their boundary conditions [1]. Therefore all wall composition with different openings could be determined as one construction element of full length with reduced shear strength and stiffness [2]. For evaluation tendency of shear strength and stiffness reduction for different fenestrations development of exact mathematical model is needed to reduce number of tests. But exact numerical models are reliable only if they passed the appropriate experimental validation and verification. Therefore, at University of Ljubljana, the series of racking tests of fenestrated cross-laminated (X-lam) solid wooden Fig.1: Example of massive wooden panels were carried out to supply experimental data needed wall with openings and principle of design with only full wall segments. for computational model development [3]. The main objective of testing was to understand the global response of fenestrated panels and to obtain data for verification and validation of response of the panel. Numerical model was developed and parametric study carried out by commercial software SAP2000 - Version 8 Nonlinear. Main
Sugiyama in M. Regarding the shape of the opening. higher value of r is associated with vertically oriented fenestration. The proposed panel area ratio r takes into account both size and shape of openings. Ratio of shear stiffness is defined as relation between shear stiffness of the wall with openings and shear stiffness of the wall without 106 . the value of r is inversely proportional to the opening area. They defined the sheathing area ratio. [6]. Regarding the size aspect. Models were systematically numerically tested by SAP2000 and results of calculation presented in form of parametric study.1). Yasumura Hideo Sugiyama and Motoi Yasumura in 1980's reported results of racking tests on plywoodsheathed timber frame specimens of reduced scale 1:3 with various openings [4] and full size walls [5] constructed according to standard specification of North American wooden frame structures.1 Research done by H. Study of fenestrated X-lam wooden walls followed the concept of research that was done on timber frame (TF) walls presented in Chapter 2. 2006 parameters of interest were related to nonlinear behaviour of anchors and elastic behaviour of Xlam solid wooden wall segments around openings. are presented in Figures 3 and 4. Parametric study resulted in diagrams which show the relation between panel area ratio and ratio of racking load and stiffness of fenestrated X-lam walls against non-fenestrated one. r (eq.Influence of openings in timber frame walls In this study the influence of openings on X-lam solid wooden wall was the main interest of research. L1 A1 L2 L L3 The ratios between shear strength F and shear stiffness K of fenestrated TF wall and non-fenestrated one.2). The sheathing area ratio was determined by: a) the ration of the area of openings to the area of wall and b) the length of wall with full height sheathing to the total length of wall. [5]. The main outcome of their research was definition of two empirical equations (eq. As far as we know there are no published research results concerning behaviour of X-lam fenestrated panels. 2. in order to classify walls based on the amount of openings a wall contains (Fig.3).International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.2) which make possible determination of shear strength and stiffness of various fenestrated timber frame walls if shear characteristics of full wall of the same dimension is known. Developed and justified mathematical model of experimentally tested panel was used for development of further mathematical models of panels having different patterns of fenestrations. Variables in equation 1 mean: r H L ∑ Li ∑ Ai panel area ratio height of the wall element length of the wall element length of full height wall segments sum of openings A2 H r= α 1+ β 1 = H ∑ Li H ∑ Li + ∑ Ai (1) α= ∑ Ai HL ∑ Li β= L ratio of openings in wall element ratio of full wall segments.2: Fenestrated timber-frame wall with fastened sheathings. With using the same definition of parameter r it is proposed herein to be named "panel area ratio" of fenestrated X-lam wooden wall (Fig. However. as obtained from tests. State of the art . In the case of non-fenestrated wall the value of r is equal to 1. For that reason all composed parts of X-lam walls were separately tested in the scope of accompany experimental programs to obtain their exact mechanical properties. there are published research results on fenestrated TF constructions [4]. Portugal November. Fig. 2.
Results obtained on full size TF walls shown slightly higher shear capacity especially at smaller ratio of openings.4). Ratio of shear stiffness was evaluated at shear deformation of 0. Portugal November. Fig.28 (3) Significant differences between monotonous and cyclic response of shear walls dictated evaluation of seismic response of a structure on mechanical characteristics of walls obtained by cyclic tests. Johnson In 1996 J. Yasumura Sugiyamova krivulja 1 Rezultati preiskav Dolan. F Ratio strižne of shear strength.6 K 0.8 1 0 0 0. C.4 0. K F Delež nosilnosti.8 1 Panel area odprtin. ratio of shear strength presents relation between shear strength of fenestrated and non-fenestrated wall. Higher difference between monotonous and cyclic response is observable at non-fenestrated TF panels.6 0. 1 Rezultati preiskav Dolan. Johnson Sugiyama.3: Comparison of Sugiyama curve and tendency line obtained by Dolan's tests for reduction of shear stiffness of fenestrated walls. Because presented empirical equations mainly based on monotonous shear test results. D. Johnson Sugiyama. Monotonic results were compared with tendency curve obtained by Japanese research (Fig.3. 2006 openings of equal dimensions and at the same level of shear strain. In diagrams which presenting influence of openings on shear capacity of TF wall the same tendency curve was determined for evaluation of shear strength ratio and shear stiffness ration. 1003 107 . F = 1.6 0. Comparison of shear stiffness ratio with panel area ratio linear relation was recognized as set in Equation 3. Analogically.2 0.2 0. Dolan and A. r r Koeficient Panel area odprtin.2 0 0 0. The experimentally obtained empirical equations are expressed by panel area ratio as set in Equation 2: K= r 3 − 2r and F= r 3 − 2r (2) Where means: F ratio of shear strength of fenestrated and non-fenestrated TF wall K ratio of shear stiffness of fenestrated and non-fenestrated TF wall r panel area ratio or sheathing area ratio as named in TF walls research 2.2 0.4 0. Delež strižne togosti. Yasumura Sugiyamova krivulja Ratio of shear stiffness.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. With decreasing of panel area ratio r or with increasing the ratio of openings α the difference becomes smaller. ratio.2 Research done by J. Shear resistance determined by cyclic tests was up to 30% less than resistance obtained by monotonous loading. C. ratio. Johnson published monotonous and cyclic test results on full size TF walls with various openings.27 r − 0. F 0.8 0. Daniel Dolan and A.4 0.01 rad of TF wall and is almost the same value also at lower shear deformations. further investigation was done on full sized TF walls loaded by cyclic horizontal loading defined according to sequential phased displacement protocol. r r Koeficient Fig. Fig. Shear stiffness was evaluated at shear deformation which corresponds to 40% of ultimate racking load.8 0.6 0.4: Comparison of Dolan's test results and Sugiyama empirical curve for reduction shear strength of fenestrated TF walls.4 K 0.
Specimens of Wall 14 were without openings while the specimens of Wall 13 had door and window opening (Fig.4 cm were produced by Austrian company KLH Massivholz GmbH as solid elements composed of three layers of cross glued lamellate (X-lam) wood. controlled horizontal load H 3-layer KLH element (t=94mm) corner connector h=105 mm with rib fixed to KLH plate by 10 Kammnails 4. The tests have been preformed at the Laboratory of the Faculty of Geodesy and Civil Engineering in Ljubljana.0/40 mm and to concrete base by 2 bolts M12 uniform vertical load 15 kN/m' Fig.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. force-displ. In presented tests four BMF corner connectors of height of 105 mm with ribs were used and placed on marked location. 108 . Slovenia using the test set up specially designed and constructed for testing of panels under different boundary conditions and constant vertical loading [1]. Fig.5). The wall specimens of dimension 320/272/9. Boundary corner connectors were placed 10 cm from the edge of the wall elements. For fixing corner connector to KLH wall plate 10 annularly nails 4. 0/40 mm was used. 2006 3. 80 Horizontal Force [kN] 60 80 60 Horizontal Force [kN] 40 20 -50 -40 -30 -20 0 -10 -20 0 -40 -60 -80 Horizontal displacement at the top [mm] W14c_C_V1/1 10 20 30 40 50 40 20 0 -50 -40 -30 -20 -10 -20 -40 -60 Horizontal displacement at the top [mm] 0 10 20 30 40 50 W13c_C_V1/1 80 Horizontal Force [kN] Horizontal Force [kN] 80 60 40 20 -50 -40 -30 -20 0 -10 -20 0 -40 -60 -80 Horizontal displacement at the top [mm] W14c_C_V1/2 10 20 30 40 50 60 40 20 0 -50 -40 -30 -20 -10 -20 -40 -60 Horizontal displacement at the top [mm] 0 10 20 30 40 50 W13c_C_V1/2 Fig.0/40 mm and to concrete base by 2 bolts M12 uniform vertical load 15 kN/m' force-displ. For fixing it to reinforced concrete foundation beam two steel bolts M12 were used. All together four cyclic tests were performed to obtain hysteretic response of fenestrated and nonfenestrated X-lam walls (Fig.5: Configuration of X-lam Wall 13 with door and window opening and Wall 14 without openings.6: Hysteretic response of two tested X-lam walls with window and door opening (r=0. Experimental research of X-lam solid wooden walls with openings 3. controlled horizontal load H 3-layer KLH element (t=94mm) corner connector h=105 mm with rib fixed to KLH plate by 10 Kammnails 4. EN 12512 was followed for determination of cyclic loading protocol.6 and Fig. Portugal November.7: Hysteretic response of two tested nonfenestrated X-lam wall panels. [2].1 Racking tests of X-lam wooden walls with openings Two configurations of walls with equal dimensions represented by two specimens each have been tested under the same boundary conditions.41).7).
Portugal November. The main load-bearing direction of 3-layer X-lam wooden plate is defined with direction of wooden 1005 109 . Dissipation of energy presented by wideness of hysteretic loops comes mostly from deformation of corner connectors and fasteners which connect them to the wall and foundation. The moisture of these technically dried strips glued together is 12% (±2) which disables any pests. On Wall 14 anchors were fixed on the same Fig.5). The second stage is recognized in lower stiffness and by very wide hysteretic loops. while the magnitude of vertical load prevent tension deformations of anchors. Cross lamination of the timber strips has many advantages.8). fungi or insects attacks.9: Tests on X-lam wall bearing direction and 34 mm in segments for determination perpendicular direction (Fig. 9). Therefore some accompanying tests were done on composition parts to obtain their mechanical properties which were used in modelling process. considerably increases static strength and shape retention properties and enables load transfer across the entire plane of panel [7].4 cm. 2006 In hysteretic wall response two significant stages can be observed as differences in stiffness inclination and in wideness of hysteretic loops.41) reduce shear stiffness -20 W13c_C_V1/1 while shear strength remain almost -40 W13c_C_V1/2 of the same value. Strips thickness is in the case of 3-layer X-lam wall 30 mm in the main loadFig.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.2 Accompanying experimental programs for determination of mechanical properties of constituent elements The response of the tested wooden walls does not depend only on the boundary conditions and the magnitude of vertical load but mostly on the configuration and mechanical properties of the constituent elements and the assembly as a whole. 80 The results of all four cyclic tests are summarized in diagrams where W13 W14 60 envelops of experimentally 40 obtained hysteretic response of Wall 13 and Wall 14 are presented 20 (Fig. opening dictated their positions. visible as more stiff response with very tight hysteretic loops. Higher horizontal loads caused tension forces in anchors and therefore up-lift of the wall. 3. Vodoravna sila [kN] Horizontal force [kN] 3. Xlam wooden plate consists of strips of spruce stacked on top of each other and glued together forming large-sized solid crosslaminated boards. These differences are more visible at nonfenestrated wall response compared to fenestrated one. represent shear wall response where mainly shear deformations of X-lam wood material occurred with some initial contact slips between assembled elements. The responses in W14c_C_V1/1 both directions of non-fenestrated -60 W14c_C_V1/2 wall are not the same because of -80 unsymmetrical distribution of Horizontal displacement at the top [mm] Pomik na vrhu stene [mm] anchors (Fig.2. mechanical properties.8: Envelops of hysteretic response of tested X-lam walls position as on Wall 13 where door of dimension 320/272/9.4 cm. The first stage.1 Modulus of elasticity of X-lam wooden wall segments Modulus of elasticity in both orthogonal directions were determinate on the wall segments of dimension 30/30/9. It minimizes swelling and shrinkage in the board plane. It can be observed in 0 compared tests that openings -50 -40 -30 -20 -10 0 10 20 30 40 50 (r=0.
1 F max 0.3 0. In Table 2 experimentally obtained E-moduli were compared with calculated ones. 800 700 600 Sila [kN] Force [kN] 500 400 300 F2 200 100 F1 0 0 u1 0.2 0. Rough estimation of E-moduli and strengths for solid wood panels with cross layers could also be that perpendicular layers are not taken into account.001 0. 30 25 Napetost [MPa] Stress [MPa] 20 15 10 5 0 0 0.12) tg α.10: Measurements of X-lam segments during Fig.7 0.006 P1v P2v P3v P1h P2h P3h Fig. V2.002 0.4 F max V1 V2 V3 V4 average.90). 110 . Portugal November. Considered calculation of effective mechanical properties for solid wood panels with cross layers was proposed by Blass [8]. Good agreement is observed where calculated values are a little smaller and therefore on safety side.003 Deformacija Strain 0.11: An example of measured displacements compression tests for determination of E-modulus for determination of E-modulus in the main load bearing direction of 3 layer X-lam wooden wall. E-module represent on diagrams (Fig. 0 while E-modulus in perpendicular direction was signed as Ep.6 0.12: Strain-stress relation at compression loading in both orthogonal directions.011 mm/s and 0. E-modulus in load-bearing direction (Ep. where stiffness of cross layer is also taken into account.90. 2006 fibres in outer layers. The analytical approach determines homogenisation of cross section (as all wooden fibbers are oriented just in one direction) with compositions factors.4 0. Because two layers of wooden strips are oriented in the main load-bearing direction with common cross section almost double against perpendicular one. where α is the angle between linear inclination part of stress-strain relations and abscissa. Deformations in each layer were measured by deformeters (V1. Both E-moduli in plane of 3-layer X-lam wooden plate were defined according to EN 789 on three specimens for each orthogonal direction. Loading was defined by increasing of hydraulic actuator displacement. The rates were 0.9 Measured displacement [mm] Pomik [mm] Fig. Therefore the E-modulus in load-bearing direction was signed as Ep.5 0.004 0. V3 and V4) with measuring length of 200 mm. Table 1: Experimentally obtained results of Emodulus of 3-layer X-lam wooden plate in loadbearing direction and in perpendicular direction. also the ratio of E-module in orthogonal directions should be of the same value.014 mm/s what correspond to loading rates that failures occurred within time interval of 300±120 s according to standard requirements. The values of experimentally obtained E-module are presented in Table 1.8 0.1 u2 0. This simple estimation is presented on Figure 13 and can significantly underestimate mechanical properties of X-lam wooden panel. displ. in plane for both orthogonal directions.0) is about 9 GPa and is doubled regarding to E-modulus in perpendicular direction (Ep. Povpre čni pomik 0.005 0.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.
4 0.14). Loading rate with increasing displacement of hydraulic actuator was 0. P Δy Fig. while horizontal deformation were measured by two LVDT's (I1 and I2) with measuring length of 345 mm. where three specimens (P1s.9 1 P1s P2s P3s Normalized Normirana shear strižnastress napetost Fig.3 0.2 Shear modulus Shear modulus was obtained by experimental tests on X-lam wall segments. 2006 Table 2: Comparison of calculated [8] and experimentally obtained E-modulus of X-lam wooden plate in two perpendicular directions.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. At this rate of loading failure occurred within time interval of 300±120 s.8 0.2. 3. t Δx/2 P l d Δx. 1400 Strižni modul [MPa] Shear Modulus [MPa] 1200 1000 800 600 400 200 0 0 0. d α P l. shear modulus of specimen. shear strain in specimen.13: Approximate stress distribution under bending-axial loading of X-lam wooden plate.14: Diagonal test for determining shear modulus. P3s) were loaded by compression force in diagonal direction (Fig. diagonal length of specimen and horizontal in vertical displacement. width and thickness of specimen. acting force on specimen.5 0. 1007 111 . full Fig. Vertical deformations of specimens were measured by two deformeters (V1 and V2) with measuring length of 375 mm. Portugal November.1 0. Proper boundary conditions for applying diagonal load were considered by appropriate set-up shoe dimension. Where symbols mean (Fig. Δy Fig.16: Meaning symbols for shear modulus determination.6 0.012 mm/s.7 0.2 0.16): τd γd Gd shear stress in specimen.15: Shear modulus in relation to normalized shear stress. P2s.
17: Semi-cyclic up-lift corner connector test to obtain global hysteretic behaviour of anchor at tension loading.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.V.57 kg/m3 with standard deviation of 11. shear strain and shear modulus were set by following equations (Eq.C3 Displacement cyclic loading full cyclic protocol Displacement -25 -20 -15 -10 -5 0 5 10 15 20 25 30 Shear slide [mm] Fig.03. Determined mean density was 417. At higher load values shear stress distribution in specimen is not infected by local effects and therefore shear modulus is quite stable and slightly follows the curve from minimum value of 0. 16 12 8 Shear force [kN] 4 0 -4 -8 -12 -16 -30 Time Time KLH1 .5GPa was taken into account in numerical analyses of racking response of X-lam walls. 3. 16 14 12 Tension Force [kN] 10 8 6 4 2 0 -2 -4 0 Time Time KLH1-C Displacement Dispalcement semi-cyclic loading Semi-cyc protocol 2 4 6 8 10 12 14 16 18 Up-lift [mm] Fig.2.3 Tension and shear behaviour of anchors The main purpose of tension and shear tests of anchors was to determine exact global response of constituent part of the shear wall which importantly influences on its racking behaviour.65GPa and again decreases to value of 0. of 0. Applying of concentrated load at the corners of specimen doesn't assure proper stress distribution at smaller load values.O.5GPa.12 kg/m3 and with C. 3. Portugal November.3): τd = P .3 Density of 3-layer X-lam wooden segment Before testing all specimens were measured and weighed at relative humidity of 12%. 2006 On the base of measurements the shear stress. The minimum obtained value of G-modulus of 0.5GPa up to 0. 2d tan α Gd = τd γd (3) From diagram on Figure 15 it is obvious that measured deformations at initial stage of shear tests don't express correct value of shear modulus. 2 ⋅l ⋅t γd = Δx + Δy 1 (tan α + ). 112 .18: Cyclic slide connector test to obtain global hysteretic behaviour of anchor at shear loading.
dir. Material characteristics for the membrane elements of thickness of 9.8). The idea of proposed attempt was developed as much is possible exact mathematical model (Fig. 4 1 2 3 Non-linear springs 4 Fig. 4. Experimental results of racking tests as presented in upper part of the paper were the source for its verification (Fig. Numeric analysis An attempt to prediction of racking behaviour using one of commercially available software is carried out. The exact models are reliable only if they passed the appropriate experimental validation and verification. shear load was applied. In the first scope of tests. dir. Uniformly distributed vertical load Contact element . respectively. The model is consisted from orthotropic membrane elements and longitudinal sprigs which simulate anchors behaviour. named "slide tests" (Fig. tension load on anchor was applied while in the second scope of tests. one test with monotonous load and three with cyclic load.19) with taking into account realistic mechanical properties of all composed elements. characteristics of spring elements simulating contact with RC foundation and characteristics of non-linear behaviour of anchors in two perpendicular directions. For fastening corner connector to X-lam wall segment 10 annularly nails 4. named "up-lift tests" (Fig.25 ρ = 417 kg/m3 2 Non-linear spring for anchor in vert.4 cm have been defined according the results of tests on wall segments. 3 Non-linear spring for anchor in hor. Anchors of shear wall are mostly loaded in combination by tension and shear force. Portugal November.18).17). After this stage friction springs have constant load-bearing capacity and resist sliding of panel in combination with non113 1009 .7. Therefore the segments of X-lam wall were commonly fixed to foundation and tested in two perpendicular directions. 2006 In presented tests X-lam wall segments were fix to steel plate by BMF corner connectors of height of 105 mm with ribs. Springs are absolutely stiff until the shear flow in contact zone doesn't attained estimated friction force.19). The value of friction coefficient between the rough concrete and X-lam wooden surface was estimated on 0.19: Mathematical model in SAP2000 for Specimen 2 with material characteristics of FE. At cyclic tests envelops on hysteretic responses were set and basic mechanical characteristics were defined for mathematical modelling (Fig. The SAP2000 was chosen as one of widely popular program.0/40 mm were used and for fixing corner connector to steel plate two bolts M12 were used. Monotonous and cyclic (semi-cyclic) tests were performed according to standard EN 26891 and EN 12512.RC beam to KLH wall Material characteristics of plane FE: E1 = Eh = 445 kN/cm2 E2 = Ev = 900 kN/cm2 G = 50 kN/cm2 ν = 0. Contact between wall elements and RC foundation is represented by set of springs which are absolutely stiff in the direction of foundation and allow free movement away from foundation.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Three cross-lamellate glued layers of KLH walls were taken into account as homogeneous orthotropic material. Friction between foundation and wall element is modelled using bi-linear link elements placed in horizontal direction along the whole length of the lower edge of the panel. All together four tests in each loading direction were carried out.RC beam to KLH wall FH Plane FE 1 Friction element .
Models were numerically tested by SAP2000 while results of calculation are presented as response diagrams. For evaluation tendency of shear strength and stiffness reduction for different fenestrations an exact mathematical model is needed to reduce number of tests. Verified mathematical model of experimentally tested panel was used for development of further mathematical models having different patterns of fenestrations. 2006 linear springs which represent shear behaviour of corner connectors. In Table 3 the matrix of fenestrated models having wall length of 320 cm is presented.41. Namely mathematical model by using orthotropic membrane elements is not able to recognize non-linear mechanical properties of wooden cross-section. The characteristic strength of homogenised 3-layer X-lam cross section was determined by analytical approach using compositions factors [8]. All together 36 fenestrated models were set with three different lengths. 320 cm and 400 cm. Fig. Portugal November. 80 60 Vodorvana sila [kN] Horizontal force [kN] 40 20 0 -40 -30 -20 -10 -20 -40 -60 -80 Pomik na vrhu stene [mm] Horizontal displacement at the top [mm] 0 10 20 30 W13c_C_V1_1 W13c_C_V1_2 SAP SAP calculation 40 80 60 Horizontal [kN] Vodoravnaforce sila [kN] 40 20 0 -40 -30 -20 -10 -20 -40 -60 -80 Pomik na vrhu stene [mm] Horizontal displacement at the top [mm] 0 10 20 30 W14c_C_V1_1 W14c_C_V1_2 SAP calculation 40 Fig. which is parallel to 114 .20 and 21).21: Comparison of experimentally obtained and calculated response of non-fenestrated wall with unsymmetrical distribution of anchors. With comparing envelopes of experimentally obtained responses and calculated ones relatively good agreement can be observed.17 and 18). The lines are the secant of the skeleton curve defined by points at 10 % and 40% of horizontal load-bearing capacity and tangent on the upper part of the envelope. With above presented model parametric study was done to obtain influence of different openings in Xlam wooden walls on their shear capacities. and horizontal load has been applied once in positive and once in negative horizontal direction as the response of the wooden panel is not equal in two different directions as a result of unsymmetrical disposition of the anchors and openings. On Figures 20 and 21 calculated and experimentally obtained racking responses of fenestrated and non-fenestrated X-lam wooden walls are presented.20: Comparison of experimentally obtained and calculated response of fenestrated wall with panel area ratio of 0.22). In every model analysis results for each loading step were compiled and presented in a form of response diagrams as relation between horizontal force and horizontal displacement (Fig. Shear stiffness was set as an inclination line which goes through the yielding point defined according to CEN II definition. Non-linear tension and shear behaviours of anchors were determined on the base of test results (Fig. stress distributions around openings were analysed to observe compression or tension failure of cross laminated wood. respectively. (Fig. 5. Non-linear static pushover analysis using SAP 2000 has been run and the obtained results have been presented in a form of horizontal force-displacement diagrams. On Figure 22 definition of shear stiffness of calculated racking response is presented. Numerical analyses were performed as non-linear static pushover analyses. In this definition the yield limit state is set as the point of intersection between two lines. At higher loadings.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Parametric study of influence of openings The load-bearing capacity and stiffness of fenestrated wooden walls is influenced by dimensions and layouts of openings. Numerically analysed models were defined in three matrixes with wall lengths of 240 cm. For each length an additional non-fenestrated model was set and numerically tested to get ratio of shear capacities against fenestrated ones. In mathematical model vertical load of 15kN/m' has been defined. Experimentally obtained envelopes were incorporated in the model as multi linear springs. All fenestrated models have symmetrical opening which systematically increasing in length and height according to the matrix by one quarter of wall dimension.
: Dim.5 270 202.35 cm.25 10 100 100 320 100 10 120 80 Delež višine odprtine Ratio of opening height 2 4 h b 270 135 67.75 10 10 100 100 320 40 67.75 10 67.5 270 202.5 101. The shear force values at horizontal displacement of h/200 or 0.5 101.25 270 67.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.75 10 100 100 320 120 80 120 67.25 10 100 100 320 120 67.5 3 4 h d 270 202. Therefore in calculated racking response shear strength was set at deformation value of 0.5 10 100 100 320 100 10 10 100 100 320 100 10 10 100 100 320 100 10 120 80 120 33. Table 3: Matrix of mathematical models with length of 320 cm by systematically increasing opening length and opening height by one quarter of wall dimension. 1011 115 .5 80 160 80 67.75 40 240 40 33.5 101.5% of story drift were set as horizontal load-bearing capacity of calculated response.5 80 100 10 10 100 100 320 160 80 100 33.25 100 320 240 40 67.5 240 40 100 33. Portugal November.25 80 2 320 160 80 101.5 270 202. 2006 the secant through the skeleton curve at 40% and 90% of horizontal load-bearing capacity.5 270 135 67.75 80 160 80 33. According to presented matrix and opening definition also mathematical models with length of 240 cm and 400 cm were numerically analysed.5 10 100 100 100 10 10 60 180 60 10 10 20 260 20 10 The accuracy of calculated response of fenestrated model with large opening at higher stage of deformations is questionable while failure usually occurs by smashing or tearing wood fibres in corners around openings. what correspond to horizontal displacement of 1.5 40 100 10 10 100 270 40 3 320 240 40 101.stene: Delež dolžine odprtine Ratio of opening length 1 4 l = 320 cm h = 270 cm 120 l 2 4l 3 4 l 1 320 80 120 101.5 270 202.5% of story drift.75 3 4 h c 270 202.5 270 135 67.25 67.5 33.5 100 10 1 4 h a 270 67. Wall dim.
Portugal November. 320_vrata door win. On the base of observed relation tendencies two empirical formulas were defined (Eq.4 0. 400_vrata door 15 door 320_vrata win.22: Relation between calculated shear stiffness and ratio of opening area for different X-lam walls. 15 20 25 Pomik na vrhu stene [mm] Horizontal displacement at the top [mm] h/200 30 35 Fig.24 and 25) which show the relation between panel area ratio and normalized values of shear capacities expressed as ratio of racking load and stiffness of fenestrated X-lam walls against non-fenestrated one. 400_okna win.23: Relation between calculated shear strength and ratio of opening area for different X-lam walls. Fig.23) and shear strength (Fig.22: Calculated responses of three differently fenestrated and one non-fenestrated walls with length of 320 cm.2 0.5% of story drift are presented. win. win. 240_vrata door win. Strižna togost [kN/mm] Shear stiffness [kN/mm] 20 240_vrata door win.24) and ratio of opening area for different X-lam walls (wall length and shape of opening) are presented. 116 . 2006 As the normal stresses in analysed specimens with the largest opening were observed at failure value when story drift was higher than 0. For non-fenestrated wall the shear stiffness definition and shear strength evaluation at horizontal displacement correspond to 0.1Fmax 320_b1 320_b2 320_b3 320_polni full w. 320_okna 90 80 Strižnastrength nosilnost[kN] [kN] Shear 70 60 50 40 30 20 10 win.9Fmax 20 10 0 0 5 10 0.4).2 0. 240_okna win.6 0 Delež površine odprtine (a ) Delež površine odprtine (a ) Ratio of opening area [α] 0. 70 Vodoravna sila [kN] [kN] Horizontal force 60 50 40 30 0.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. The aim of preformed parametric study was to obtain regression of shear capacities with increasing of fenestrated area in X-lam solid wooden walls. door 400_vrata 10 5 0 0 0 Ratio of opening area [α] 0.4Fmax 320_b stiffness strength 0.6 Fig. Parametric study resulted in diagrams (Fig. it can be conclude that numerical models give good results up to this deformation stage. 25 240_okna win. 320_okna win.4 0. 400_okna win. When stress level corresponds to wood failure in crosslaminated wood the rest part of calculated racking response is presented by dashed curve. On the following diagrams relationship between calculated shear stiffness (Fig.5%.
Fig. Delež strižnestiffness. rr 0 0 0. r Fig. Panel area ratio.8 1 0. Ratio shear strength.4 0.8 0. Panel area ratio.2 0.6 0. K opening = K full ⋅ r 2−r and Fopening = F full ⋅ r (2 − r ) (4) If simple formulas will be confirmed by further experimental test results. they could effectively serve in X-lam timber structure design. Panel area ratio.4 0. stiffness.2 0 0 0.8 Koeficient odprtin.2 0 0 0. 1013 117 . 2006 1 Delež K K Ratio of strižne shear togosti. The experimentally obtained empirical equations (Eq.4 0. r. Portugal November. KK Ratio of shear 0. Fig.26) while for reduction of shear strength tendency curve with opposite curvature was defined.2 0. togosti.4) are expressed by panel area ratio and could be use for direct evaluation of reduced shear capacities if shear stiffness and shear strength of non-fenestrated wall are already known.8 1 Koeficient odprtin.2 0.8 0.6 0. r.2 0.6 0.25: Calculated shear strength reduction regarding the panel area ratio for X-lam walls.4 Dolžina stene: 240 cm Wall length: : length Dolžina stene: 320 cm Wall length: : length Wall length: : Dolžina stene: 400 cm length Experimental resultrezultat Eksperimentalni 0 0.4 0.6 0. r r 0 Panel area ratio. rr 1 Fig.4 0.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.6 F = r (2 − r ) 0.4 0.27: Comparison of shear strength reduction for TF wall and X-lam solid wooden walls regarding the panel area ratio.24: Calculated shear stiffness reduction regarding the panel area ratio for X-lam walls. Comparing influence of openings on shear capacities of X-lam and TF walls similar tendency curve was obtain for reduction of shear stiffness (Fig.2 r K= 2−r 0. Delež of strižne nosilnosti. r Koeficient odprtin. FF Ratio shear strength.8 1 Koeficient odprtin. where horizontal building resistance have to be analysed.8 Okvirne Timber stene frame walls X-lam Lepljene KLH walls masivne stene 0.2 0.8 Dolžina stene: 240 cm Wall length: : length Dolžina stene: 320 cm Wall length: : length Dolžina stene: 400 cm Wall length: : length Experimental result Eksprimentalni rezultat 1 0.6 0.6 0.4 0.FF 0. 1 Okvirne Timber stene frame walls X-lam Lepljene KLH walls masivne stene 1 Delež of strižne nosilnosti.6 0.26: Comparison of shear stiffness reduction for TF wall and X-lam solid wooden walls regarding the panel area ratio.
C. and Zarnic.. Finland. 2001. Lahti. UL FGG. Ljubljana. because failures are mostly concentrated in anchoring areas and in corners around openings with smashing and tearing of wood. P. Conclusion Non-fenestrated cross-laminated wooden walls have relatively high stiffness and load-bearing capacity.D. Portland. Ljubljana. Aicher.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. S. Oregon. Johnson. J. Dolan. Portugal November. paper 19-15-3. Proceedings of the 8th World Conference on Timber Engineering. WCTE 2004. M. M. 2004.. 8p. Preformed experimental research and parametric study resulted in diagrams that could serve for engineering judgment of fenestration influence on stiffness and load-bearing capacity of X-lam solid wooden walls. Fellmoser. Slovenia.. It is obvious. "Racking Resistance of Wooden Frame Walls with Various Openings" Proceedings CIB-W18. Slovenia. Massivholz GmbH KLH Technical Documentation Blass. "Performance of Long Shear Walls with Openings" International Wood Engineering Conference. "Influence of openings on Shear Capacity of Wooden Walls". H. 88-98.. "Testing of Wooden Wall Panels Applying Realistic Boundary Conditions". but load-bearing capacity is not reduced so much. already started at University in Ljubljana. 1986.. A. "Design of solid wood panels with cross layers". Therefore. UL FGG. New Orleans. R. will yield in the verification herein presented empirical equations. For evaluation tendencies of shear strength and stiffness reduction for different fenestrations development of exact mathematical model was developed to reduce number of tests. pp. Dujic. 95p. "Experimental Supported Modelling of Response of the Timber-Framed Wall Panels to Horizontal Cyclic Load".. [1] References Dujic. 2006 6. D. Furthermore widely fenestrated X-lam panel having lower shear stiffness.. openings in surface equal up to 30% of entire wall surface do not much influence on load-bearing capacity when stiffness is reduced for about 50%. Proceedings of the 9th World Conference on Timber Engineering WCTE 2006. Science and Sport of the Republic of Slovenia. and Sugiyama. "Shear Properties of Plywood-sheathed Wall Panels with Opening" Trans. 2006. USA. For evaluation influence of shape and area of an opening the same definition of coefficient of openings was used as in timber frame constructions. S. 239p. 24p. pp. Ph. Diploma Thesis (in Slovenian). April 1984. pp. following the racking test results.. B. B. J. H. Louisiana: 1996. 2001. the critical elements that govern the wooden cantilever response to earthquake excitations are anchors connecting panels with building foundation.. Additional experimental tests and numerical analyses will enlarge the knowledge related to lateral stiffness and stability of X-lam wooden walls with openings. Faculty of Civil and Geodetic Engineering. Thesis (in Slovenian). 543-548. Florence. 7. [2] [3] [4] [5] [6] [7] [8] 118 . Italy. Yasumura. June 14-17. of the Architectural Institute of Japan. No. Yasumura. August 6-10. Klobcar. The load bearing capacity and stiffness of fenestrated wooden walls is influenced by dimensions and layout of openings. 338. Both financial contributions are gratefully acknowledged. Acknowledgment Presented research program was carried out partly in the scope of industrial project funded by the Austrian Company KLH Massivholz GmbH and in the scope applied research project named "Upgrade of Earthquake Resistance of Timber Structures" which was supported by Ministry of Education.. The aim of the experimental research and parametric study presented herein is to provide the first information about tendency of reduction of racking strength and stiffness of X-lam wooden walls with openings. 337-344. 8. Additional tests.
International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. timber buildings with various constructive systems are becoming day by day a stronger and economically valid alternative to their counteracts built with concrete and masonry. The connections have been optimised and the lateral and cyclic resistance established. seismic performance. different wall configurations have been considered taking into account the influence of the anchoring system. pseudo-dynamic tests. Italy and coordinated by the CNR-IVALSA (Italian National Research Council – Trees and Timber Institute) investigating in all aspects of the building behaviour: mechanical properties of the material. their diffusion is increasing. durability and. no recommendations are given regarding this constructive system. 1) are gaining a broader acceptance. where until now timber structures had been less common within the residential housing market. This paper presents the results of quasi-static in-plane cyclic tests on XLAM walls and of pseudo-dynamic tests on a full-size building. For the quasi-static tests. 75 38010 San Michele all'Adige (TN) Civil Engineer. graduated 1994 at Florence University. a comprehensive testing programme has been undertaken. A final push-over test was carried out. especially buildings made of crosslaminated wooden panels (XLAM – Fig.it IVALSA-CNR Via Biasi. the opening layout. building physics.cnr. fire. Three different configurations differing in the opening layout have been tested applying two different earthquakes (El Centro and Kobe) at different peak ground accelerations (PGA of 0. Fig. Neither for construction details nor for the seismic behaviour factor to be used in seismic design of this new typology of wooden buildings. graduated 2004 at Technical University of Technology in Karlsruhe. acoustics. 1: XLAM panel The goal of the seismic part of this research project is to undertake a comprehensive investigation in order to obtain information on the overall seismic behaviour of multi-storey buildings made of cross-laminated wooden panels. consisting in the following stages: 119 .5g). Even in seismic prone zones like the majority of the Italian territory. 2006 Quasi-Static and Pseudo-Dynamic Tests on XLAM Walls and Buildings Marco Pio LAURIOLA Researcher lauriola@timberengineering. the magnitude of vertical loads and the interstorey connection. The pseudo-dynamic tests were carried out on a one-storey specimen with 7x7m in plan. Even in Italy. particularly. However. running an engineering company dealing with timber constructions Civil Engineer. seismic performance 1. Germany Summary The SOFIE Project is a comprehensive research project on multi-storey XLAM buildings financed by the Trento Province.it Carmen SANDHAAS Researcher email@example.com and 0. Introduction Throughout Europe. Portugal November. Keywords: cross-laminated timber (XLAM). in Eurocode 8. any information can be found. Therefore. quasi-static tests. the European seismic code for earthquake resistance of buildings.
This system is hence a very rigid system with much less energy dissipation and ductility characteristics in comparison to other conventional wood-frame constructions such as those built with the platform frame system. Portugal November. in-plane cyclic tests on wall panels with different connections and opening layouts and with different dimensions and amounts of vertical load. All the wall specimens have been anchored using commercial anchorage systems (holdown and steel angle anchors) and only in a few cases home-made holdown anchors were used in order to obtain better overall strength and stiffness properties. i. Construction System A timber building made of cross-laminated wooden wall and floor panels is a modular system where all panels (e. 120 . • pseudo-dynamic tests on a one-storey specimen in 3 different opening layouts in the external walls parallel to the shaking direction and without vertical load. The cross-laminated panels have an overall behaviour of a massive wooden panel.e. insulation and plumbing. they are very stiff and strong. 3. mainly angle brackets with nails and/or screws.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Afterwards. 2006 tests on connections. • shaking table tests on a three-storey building of about 7m x 7m in plan and 10m of total height with a pitched roof in 3 different configurations applying different earthquakes at growing levels of PGA (jointly with NIED. all tests were designed in order to evaluate the behaviour of this type of construction for different connection layouts during seismic loading. Therefore. the single modules are then assembled connecting the panels with mechanical fasteners. is carried out. In this paper. National Institute for Earth Science and Disaster Prevention. g. 2: Construction System zones. walls with openings and floor slabs) are pre-cut in the factory using the advantages of CNC-machines.95m wide and 2.95m high XLAM panels (see Fig. • • 2. four different configurations have been tested as can be seen in Table 1 below. Quasi-Static Reversed Cyclic Tests on Wall Specimens 3. The energy dissipation and ductility is only occurring due to the connections concentred in few Fig. All panels have 5 layers of 17mm thickness each resulting in a panel thickness of 85mm – which is a standard panel for wall elements of 2-3 storey buildings. the results of the quasi-static cyclic tests and of the pseudo-dynamic tests will be presented. a higher or lower ductility of the entire system can be achieved.1 Test Layout All tests have been undertaken on a 2. A ramp test carried out to establish the performance of the material itself has confirmed this statement. On site. Till now. Japan). the interior and exterior finishing. 1 for material). Depending on the connections.
95m by 2.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Annular ringed shank nails 4x60 are used to fasten the steel connectors to the panel. 18.5kN/m (based on the 1st storey wall of a real three storey building) 50 425 1000 2950 1000 425 50 Vertical load: Configuration B Panel dimension: Connection type: 2. 2950 2815 50 650 1550 2950 650 50 135 2950 Vertical load: 10.0x116 and two holdowns.95m by 2.2kN/m (based on the 2nd storey wall of a real three storey building) Configuration C Panel dimension: Connection type: 2. It must be stated though that in this test configuration with the floor slab being on top of the wall.95m without opening.95m without opening. 50 2850 2950 50 2815 2950 135 Vertical load: 18.95m by 2.20m central door opening 1003 121 . Only the shear connection was tested. All other connections (upper connection to steel beam and floor slab connection to steel base) are designed to be stronger than the connections to be tested.20m or 1. Holdown anchors in the first test 1 (see Table 2) have been "home made". The base connection and the connection of the floor slab to the upper steel beam are designed to be stronger than the connections to be tested.80m by 2.95m with a 1. As in the test 1 the nails fixing the steel angle to the floor slab were pulled out. 2006 Table 1: Tested wall configurations Configuration A Panel dimension: Connection type: 2. two additional screws were used in test 2.95m without opening Connection to base with three BMF angles 90x48x3. included a floor slab of 135mm underneath the wall panel Inter-storey connection (floor to upper wall) with two BMF angles 105 with reinforcement and two standard holdowns HTT22 by SIMPSON BMF Annular ringed shank nails 4x60 are used to fasten the steel connectors to the panel.95m by 2. included a floor slab of 135mm on top of the wall panel Inter-storey connection (floor to lower wall) with inclined screws 8x260 and two standard holdowns HTT22 by SIMPSON BMF Annular ringed shank nails 4x60 are used to fasten the holdown to the panel. Portugal November. The tests 2 and 3 have been carried out with a standard holdown HTT22 by SIMPSON BMF.5kN/m (based on the 1st storey wall of a real three storey building) Configuration D Panel dimension: 2. The quantity of nails between tests 2 and 3 has been changed in order to control failure occurring either in the holdown anchors or in the nail connection. no uplift forces are occurring. All other connections (upper connection to steel beam and connection to steel base) are designed to be stronger than the connections to be tested.
04 mm/s.04 mm/s.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.2 10. A total of 14 tests were carried out under monotonic and reversed cyclic loads.2 10.5 18.5 10. 2006 Connection type: 50 425 400 2200 1200 2950 400 425 50 2950 Connection to base with two BMF angles 90x48x3.5 18.8x2.5 18. All other connections (upper connection to steel beam and connection to steel base) are designed to be stronger than the connections to be tested.5 18.5 18.5 18.3 Test Procedure The ramp monotonic tests were performed according to EN26891 "Timber structures – Joints made with mechanical fasteners – General principles for the determination for strength and deformation characteristics" in displacement control at a rate of 0.: A/1b = configuration A.2 10.0x116 and two standard holdowns HTT22 by SIMPSON BMF Annular ringed shank nails 4x60 are used to fasten the steel connectors to the panel.2 1. EN26891 is applied with a constant displacement control and not under load control as required in the document.2 18.5 Opening size [mxm] 1. 3 shows the test set-up with the actuator used to undertake ramp and cyclic testing.5 18.5kN/m (based on the 1st storey wall of a real three storey building) The following table 2 is summarising the test carried out. 122 .2x2. The reversed cyclic tests were performed according to EN12512 "Timber structures – Test methods – Cyclic testing of joints made with mechanical fasteners" in displacement control at a rate of 0.5 18. Table 2: Tests carried out Connections CASE Subcase 1a 1b 2a 2b 3a 3b 4b 1a 1b 2a uplift shear Vertical load [kN/m] 18.2 Test protocol monotonic cyclic monotonic cyclic monotonic cyclic cyclic monotonic cyclic monotonic cyclic monotonic cyclic cyclic monotonic Homemade with 29 nails BMF 07716 with 11 nails Homemade with 29 nails BMF 07716 with 11 nails BMF HTT22 with 29 nails BMF 07716 with 11 nails A BMF HTT22 with 29 nails BMF 07716 with 11 nails BMF HTT22 with 14 nails BMF 07716 with 11 nails BMF HTT22 with 14 nails BMF 07716 with 11 nails BMF HTT22 with 14 nails BMF 07716 with 11 nails BMF HTT22 with 14 nails BMF 7105 with 8+8 nails BMF HTT22 with 14 nails BMF 7105 with 8+8 nails BMF HTT22 with 14 nails BMF 7105 with 8+10 nails B and 2 screws 8x140 2b BMF HTT22 with 14 nails BMF 7105 with 8+10 nails and 2 screws 8x140 1a BMF HTT22 with 14 nails Inclined screws 8x260 C 1b BMF HTT22 with 14 nails Inclined screws 8x260 1b BMF HTT22 with 14 nails BMF 07716 with 16 nails D 2a BMF HTT22 with 14 nails BMF 07716 with 16 nails (later reference to tests is as follows. Fig. test 1b) 2. Vertical load: 18. Portugal November.
150 150 100 100 50 50 load [kN] 0 load [kN] test A1a test A1b 0 -50 -50 -100 -100 test A2a test A2b 100 -150 -100 -50 0 50 -150 -100 -50 0 50 100 slip [mm] slip [mm] Fig 4: Configuration A. Portugal November.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. ramp and cyclic test 1 Fig 5: Configuration A. 2006 Vertical load actuators Horizontal actuator Fig 3: Test Set-up. For sake of simplicity upper stabilizing transverse steel beams are not reported. The displacement has been cleaned from any error such as relative slip between wall specimen and the test stand. 3.2 Outcomes The following figures show the test results in terms of load-slip curves. ramp and cyclic test 2 1005 123 .
In Configuration B the first test (B/1) showed a quite big impairment of strength (about 35% as it can be observed in Fig. Portugal November. ramp and cyclic test 2 150 100 100 50 50 load [kN] 0 load [kN] test C1a test C1b 0 -50 -50 -100 -100 test D1b test D2a 100 -150 -100 -50 0 50 -150 -100 -50 0 50 100 slip [mm] slip [mm] Fig 11: Configuration D. 2006 150 150 100 100 50 50 load [kN] 0 load [kN] test A3a test A3b 0 -50 -50 -100 -100 test 4b 100 -150 -100 -50 0 50 100 -150 -100 -50 0 50 slip [mm] slip [mm] Fig 6: Configuration A. ramp and cyclic test 1 150 Fig 9: Configuration B.15).International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. a stiffer holdown connection (with a 8 mm thick steel plate instead of the 3 mm of thickness of the BMF HTT22) lead to a higher lateral load carrying capacity but a lower ultimate displacement and a less ductile behaviour. ramp and cyclic test . Therefore. ramp and cyclic test (ramp test specimen 2a has a bigger opening) The test results are summarised in Table 3. In Configuration A. in test B/2 2 two screws were inserted instead of 124 Fig 10: Configuration C. cyclic test 4 150 100 100 50 50 load [kN] 0 load [kN] test B1a test B1b 0 -50 -50 -100 -100 test B2a test B2b 100 -150 -100 -50 0 50 100 -150 -100 -50 0 50 slip [mm] slip [mm] Fig 8: Configuration B. as it was expected. 8) due to the pulling out of nails connecting the steel angle to the floor slab during cyclic test (Fig. ramp and cyclic test 3 150 Fig 7: Configuration A.
pulling-out of nails fixing the steel angle to the floor slab Yielding of nails and connectors.4) Ultimate displacement1 [mm] 44. holdowns and nails. for the test C/1 carried out.1 5441.6 (96.5% 1 B 2 75. Portugal November.0 (102. yielding of nails and steel angle Yielding of nails. 3 Equivalent viscous damping. Secant stiffness between 10% and 40% of maximum load in the monotonic test.1% 98.9 - 6744.0) 116.3 129.1% 17. 2006 the first two nails obtaining a significant reduction of the above mentioned impairment of strength.5% 91.0 5401.1 117.5 115.3 Discussion and Conclusion The layout and design of the joints is influencing strongly on the overall behaviour of the structural system.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. See [2] for reference. 3.7 - Displacement at 80% of maximum load after the maximum load is reached in the monotonic test. In Configuration C a preliminary test was carried out with 15 inclined screws which proved to be too rigid.4 (112.4 νeq3 Observed failure mode Yielding of nails until shear failure of nails in holdown connections Tension failure of holdown. local failure of timber in compression Large yielding of steel angles.6 (113. lastly tension failure of holdown Yielding of nails and connectors.8) 67.5) 117.5 - 1 D 2 1 2 - - 12.4% 13.8 85.1) 62. rolling shear failures in floor slab Yielding of steel angles. The wooden panels behaved almost completely rigid in comparison with the stiffness of the anchor-panel-anchor system.8% C 1 132.4 - 13. holdown and steel angle Yielding of nails. 13 and Fig.9 130.9) 65.6 54.5 15. Hence.3) 100.4 68.4 18.0 - 93.0% 2 A 3 4 42. slight pullingout of nails and screws fixing the angle to the floor slab Yielding of screws. holdown and steel angle.6 (83. 1007 125 .2) 86. Lateral load capacity [kN] Ramp Cyclic test test 140. All forces and displacement are concentrated on a rather small region of the panel which is then leading to local failure phenomena as shown in Fig. local failure of timber in compression 1 11.2 (62. In Configuration D it was proved that no holdowns at the opening are necessary due to the high stiffness of the panel material. 9 screws were used which gave the system the ductility necessary.8 (80. holdowns and nails.5 7260.9 Secant Stiffness2 [N/mm] 5969.1 (120.4 6789. Table 3: Test results and observed failure modes for cyclic tests CASE Sub.7 2169. 15.
it can also be 126 . The forces acting on the structure are calculated considering the weight of the structure and the acceleration values of a real earthquake ground motion record. g.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 2006 Fig 12: rigid panel with gap (D/1b) Fig 13: local failure (A/1b) Fig 15: local failure(B/1b) Fig 14: holdown and nail yielding(D/2a) Therefore. if the connections coincide with gluing errors. Of particular importance for this system is the principle of the hierarchy of resistances in order to provide ductility to the system. The velocity of the load application is reduced to the velocity of a quasi-static test. Therefore this system promises to be suitable for seismic purposes. This means. but is considered only theoretically. a pseudo-dynamic test cannot consider the real velocity of an earthquake and hence the viscous damping does not take place within the tested specimen. Pseudo-Dynamic Tests on a One-Storey Specimen Pseudo-dynamic tests are a hybrid approach representing a "slow earthquake". As the building's mass must be considered in the algorithm calculating the applicable forces at each step. this mass does not have to be a "physical" additional mass on the building. 4. It has a high stiffness but still good ductile and dissipating performances. Nevertheless the hysteresis loops showed an equivalent viscous damping of 14% as average (12% considering only tests on the base connections). it is confirmed that all the dissipated energy is resulting from the connections (local problems are highlighted by this system: e. Portugal November. material failures do occur). The actual stiffness values obtained from one step are taken into account when calculating the next step.
25m • Third configuration: asymmetric layout: one external opening=4. internal opening=2. 2006 purely fictitious.50g. Cyclic tests are necessary to evaluate the behaviour of shearwalls.25m • Second configuration: symmetric layout: all openings=2. Six steel angles on every 7m long shear wall are designed fixed with 11 annular ringed shanked nails. The fastening system can be seen in Fig. As 3D structures can be tested under real (though slow) earthquake loading. The one-storey specimen was 7x7m in plan and 3m high (see also Fig. A one-storey specimen in three different configurations was tested applying two earthquake records. Fig. 2 which shows the specimen in the first configuration. The window openings in the walls orthogonal to the excitation were 1.20m. the system is reduced to a SDoF-system along the actuator direction.1 Test Set-up As only one hydraulic jack is used to apply the forces. 2). 4. The holdown anchors in the corners and at the openings are designed to take the uplift forces whereas the steel angles take the shear forces. 4.25m As can be seen in Fig. El Centro and Kobe JMA. no vertical load was applied on the building – but the building's total weight of 50 tons was considered in the test control algorithm. 16.10x1.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.2 Outcomes Table 4 shows the tests carried out and their results in terms of ultimate load-slip data. a notional mass only existent in the computer algorithm. cyclic tests were carried out to establish the stiffness of the structure.20m. more realistic conclusions can be obtained with pseudo-dynamic tests in comparison with cyclic tests on wall specimens. but this does not necessarily mean that the seismic performance of the entire structure can then be sufficiently predicted. Portugal November. in two different intensities.15g and 0. 16: specimen for pseudo-dynamic tests in third configuration The three different configurations in which the pseudo-dynamic test was carried out.0m. This is necessary to evaluate a possible damage occurred during previous loading. losing the control of the DoF 'rotation of the upper floor'. The pseudo-dynamic tests were carried out at the University of Trento. other openings=2. The holdown anchors at the corners and openings feature 12 annular ringed nails each. 0. are different in terms of the openings in the walls parallel to the displacement input: • First configuration: symmetric layout: external openings=1. The stiffness 127 1009 . The fastening of the building to the steel base is designed to be stronger than the connections to be tested In between the different pseudo-dynamic tests.
26m The change in stiffness after the big earthquakes with 0.3 21.8 10.54 embedding of nails in angles.6 94.6 352.50g PGA test. i.5 cyclic test kobe 0. steel angles and holdowns replaced) 27. This initial stiffness is then reduced to 29kN/mm after the Kobe 0. 22.17 no repair of previously damaged house undertaken house repaired (screws.50g PGA (Fig.98 5. Table 4: Testing schedule and results of pseudo-dynamic tests denom.06 asymmetric opening of 4. psd 7 psd 9 psd 10 psd 12 psd 13 psd 14 psd 15 psd 16 psd 19 psd 20 psd 21 psd 22 psd 23 psd 24 psd 25 psd 26 psd 27 psd 28 psd 31 psd 32 psd 33 psd 34 psd 35 type cyclic test kobe 0.50g PGA can be clearly seen.25m) which is 27kN/mm.4 59.15 cyclic test el centro 0. Due to the missing "physical" vertical load certainly the one-storey building reached higher displacement values as it would have been the case with the real high vertical loads being actually applied. e.15 cyclic test kobe 0.47 nail failure on some holdowns. floor slabs moved on walls 2.05.15g the structure behaved nearly elastically.2 +/.56 some creaks and groans 3. It can be seen in Figs 17 to 20 that for 0. In the first configuration for example. 2006 values obtained through cyclic tests are also indicated.70 some creaks and groans ultimate force [kN] ultimate displacement [mm] comments configuration 30.20m 01.4 22.5 cyclic test cyclic test cyclic test kobe 0.3 115.5 cyclic test cyclic test cyclic test 7.4 44.15 cyclic test el centro 0. The following graphs show a typical time-displacement graph for Kobe JMA with 0. vertical joints moved. In the first configuration with the smallest openings. it can be seen that the construction system is very stiff but still ductile.8 28.1 68. A similar reduction between initial stiffness and stiffness after the pseudo-dynamic tests with 0. 25kN/mm.2 451. This means. 128 .2 43.00m).06.06 openings with 2. the initial stiffness with 47kN/mm is naturally the highest of all three configurations.94 28.0 24.15 cyclic test kobe 0.5mm.3 513.2 84.5 cyclic test el centro 0.5 cyclic test el centro 0.00m 31. remaining vertical joint opened to huge gaps (especially on the side with the big opening) house repaired (screws. that the initial stiffness of the third configuration (asymmetric big opening of 4.0 548.8 116. It thus confirms that the behaviour of the wall is due to the connections and not due to the wooden panel for low magnitudes of shear force. as in the cyclic tests. is similar to the one of the second configuration (symmetric openings of 2. It is a bit surprising though.5 25. Again. 17) and some typical load-slip data of the pseudo-dynamic tests.65 no repair of previously damaged house undertaken 43. it resisted to a base shear of 318kN reaching a displacement of H/133. Portugal November.50g PGA is observed in the second configuration where the initial stiffness of 27kN/mm is reduced to 10kN/mm.15 5.40mm +/-150mm 7. steel angles and holdowns replaced) stiffness [kN/mm] 46.8 24.0 318.05.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. holdowns seem intact. that the bigger opening on one side did not affect the building very much.06 external openings with 1.
first configuration Load-Slip 150 Fig. first configuration Load-Slip 600 Fig. Load-Slip 150 Load-Slip 150 100 100 50 50 kN kN -6 -4 -2 0 mm 2 4 6 8 0 0 -50 -50 -100 -100 -150 -8 -150 -8 -6 -4 -2 0 mm 2 4 6 8 Fig 17: Kobe JMA 0.15g. 20: El Centro 0. the graphs are not representative. 19: El Centro 0.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.15g. Portugal November. 18: Kobe JMA 0.15g. 2006 The results for El Centro with 0. second configuration Load-Slip 600 400 400 200 200 kN 0 kN -60 -40 -20 0 mm 20 40 60 80 0 -200 -200 -400 -400 -600 -80 -600 -80 -60 -40 -20 0 mm 20 40 60 80 1011 129 .50g are not shown as the house had not been repaired before testing and hence. second configuration Load-Slip 150 100 100 50 50 kN 0 kN -6 -4 -2 0 mm 2 4 6 8 0 -50 -50 -100 -100 -150 -8 -150 -8 -6 -4 -2 0 mm 2 4 6 8 Fig.15g.
5g and a very simple numerical model.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.50g. Load-Slip 600 400 200 kN 0 -200 -400 psd14 psd24 psd32 -60 -40 -20 0 mm 20 40 60 80 -600 -80 Fig.50g. 21: Kobe JMA 0. 2006 Fig. The modelling is still undergoing and the results of the modelling will be topic of a future publication.50g.50g on third configuration Fig. a simple oscillator. 22: Kobe JMA 0. first configuration time-displacement 80 CH1 CH2 CH15 displacement MTS CH27 Fig.50g. second configuration Load-Slip 600 60 400 40 200 20 mm kN 0 500 1000 1500 2000 sec 2500 3000 3500 4000 0 0 -20 -200 -40 -400 -60 -80 -600 -80 -60 -40 -20 0 mm 20 40 60 80 Fig 23: time-history graph of Kobe JMA with Fig 24: Kobe JMA 0. 26 shows an overlap in terms of force-displacement of the result for Kobe JMA with 0. 130 .50g on all three configurations The following Fig. 25 shows the overlap of the three pseudo-dynamic tests with Kobe JMA 0. 25: Overlap of Kobe JMA with 0. Portugal November. third configuration 0.
28: Detail of floor slab and wall orthogonal to load application 1013 131 . third configuration Fig. 26: Overlap of Kobe JMA with 0.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Fig. 2006 load-slip 600 400 200 kN 0 -200 -400 Pseudodynamic test Numerical model -600 -80 -60 -40 -20 0 mm 20 40 60 80 Fig.50g.50g and numerical model Following some photos to illustrate the building's behaviour and the connection failures which have all been ductile failures. Portugal November. 27: Holdown yielding after Kobe JMA 0.
Again. A special thank to the whole team dealing with the seismic investigations of this project. e. Gabriele Bonamini. Proper connections for this construction system should be developed to really utilise the strength capability inherent in the system. Alessio Bonelli. 2006 Fig. [1] References EN26891. to Ario Ceccotti. Of particular importance for this system is the principle of the hierarchy of resistances in order to provide ductility to the system. i. This construction type seems to be very promising for seismic regions. All dissipation and ductility is only depending on the connection design. Ing. 7. General Conclusions Still the gap between strength of material and strength of connections is too big.3 Discussion and Conclusion The pseudo-dynamic tests showed that the connections optimised during the cyclic test series behaved very well also in a more realistic case. 5. 30: Nail failure and yielding of steel angle after final push-over 4. Acknowledgements This paper is published in the frame-work of the SOFIE project on XLAM buildings funded by the Province of Trento. 29: Nail failure in corner holdowns after final push-over Fig. The XLAM construction system is rigid with good dissipating performance at the same time. the strength of the connections can be increased approaching the strength of the material (certainly without overtaking it) without loosing the ductility. Mario Moschi. Maurizio Follesa.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Florence University. Massimiliano Natale and Mario Pinna of CNR-IVALSA and to Saverio Giordano of Dept. 6. "Timber structures – Joints made with mechanical fasteners – General principles 132 . of Civil Engineering. Portugal November. Also Professor Motoi Yasumura gave invaluable help in preparing the pseudo-dynamic tests. Paolo Zanon and Riccardo Zandonini. the XLAM panels used to construct the walls and floor slabs proved to be very rigid. Without the help of the technicians of Trento University. the pseudo-dynamic tests would not have been possible. Marco Molinari and Dr. the professors Maurizio Piazza.
1991. EN 1998-1:2004.. "Timber structures – Test methods – Cyclic testing of joints made with mechanical fasteners". pp:V2 179-186.. Mancosu Editore. CEN Brussels. "Eurocode 8: Design of structures for earthquake resistance . 2005. 2006 [2] [3] [4] [5] [6] [7] [8] for the determination for strength and deformation characteristics".. 1991. "Le strutture di legno in zona sismica".. CIB W18 Meeting. New Orleans. CEN Brussels. EN1995-1-1:2004. 2003. Lezione C17.. Portugal November. P. CLUT. 1996. "Connessioni nelle strutture in legno soggette ad azioni sismiche". seismic actions and rules for buildings". 2004. France.II. Ceccotti A. Bordeaux.. International Wood Engineering Conference. 456pp. CEN Brussels. Follesa M. 1996 1015 133 . Torino. CEN Brussels. USA. Ceccotti A. "Quasi-Static Reversed Cyclic Testing of Nailed Joints". 2004. Lauriola M.Common rules and rules for buildings". Karacabeyli E. Ceccotti A. Part 1-1: General . Karacabeyli E. Ceccotti A. Il Manuale del Legno Strutturale Vol. EN12512.Part 1: General rules..International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Rome. "Eurocode 5 – Design of Timber Structures. "Test Results on the Lateral Resistance of Nailed Shear Walls".
International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.mk rzarnic@fgg. Research campaigns have resulted in new knowledge and approaches.D. Roko ZARNIC Ph. Professor University of Skopje Faculty of Civil Engineering Skopje.shear walls – has been investigated with appearance of new sheathing panel products. 3) in a row. demands an extensive worldwide testing campaign that would enable the comparison of behaviour of different joint configurations. The main purpose of testing and performed numerical analysis was to obtain the necessary data for modelling inelastic response of lateral resistance of timber shear walls.D. Years of experience and investigation have led to the great extent in the field of structural connectors and cleared up the minimum requirements for structural detailing in seismic areas [1]. Civ. Civ. Slovenia email@example.com. 2006 Influence of loading protocols on hysteretic response of sheathing-to-framing nailed joints in shear walls Tatjana KOCETOV MISULIC M. Slovenia . were the initial demands for several European regional research projects on this topics (Slovenia Republic of Macedonia. cyclic. [2]). The joints (timber frame-double sided particleboards-nails E28/50) were tested with different number of nails (1.D.. especially after great natural disasters. Slovenia email@example.com. Republic of Macedonia . sheathing-to-framing joints. fasteners and anchors. also the behaviour of sheathing to framing joints under different loading procedures is recognized as essential for prediction and modelling of TSW performance [3]. Key words: timber shear walls.. Serbia email@example.com). Assistant (1989) and lecturer (1998) for Timber structures in Civil engineering and Timber bearing construction in Architecture. assessment and repair of traditional and contemporary timber structures. "dynamic" and near fault) is made in order to compare joint behaviour. Eng. mechanical properties of local timber and sheathing materials. loading protocols 1. Introduction The huge number of residential and commercial timber frame buildings (TFB) around the world has raised the question of their safety and constant improvement. Eng.si Kiril GRAMATIKOV Ph. The variety of joint design.Sc. Associate Professor University of Ljubljana Faculty of Civil and Geodetic Engineering Ljubljana. Portugal November.edu. anchorage requirements. Eng.. Civ. panel proportions. etc. Assistant University of Novi Sad Faculty of Technical Sciences Novi Sad. and influenced building codes. Civ.si Summary The experimental investigation of sheathing-to-framing joints under the different loading protocols (monotonic.ns.Czech Republic. 2. different types of fasteners.uni-lj. Slovenia. Constant needs for database on sheathing/framing joints performances.ukim..yu Bruno DUJIC Ph. nails. Eng.. As the first phase of the bilateral SLO-MAK project "Development of Analytical Methods for Dynamic Analysis of TFB Seismic Resistance" the experimental investigation of joints under the 135 . Besides a number of structural detailing issues that should be satisfied in the conceptual design of timber shear walls (TSW) to provide good seismic performance (sheathing thickness. Macedonia Born 1962. chord and strut design. nailing. Teaching Assistant University of Ljubljana Faculty of Civil and Geodetic Engineering Ljubljana. as well as potential possibility of response generalisation. analytical and experimental research in timber joints and shear walls. The seismic performance of basic structural elements of TFB . All experimental testings were performed at the Faculty of Civil and Geodetic Engineering at University of Ljubljana. The earthquake resistance has always been very topical issue.
e. EN 384.5. with idea to simulate the continual sheathing to framing joint in typical TSW [4]. material: S-P-F C30 PB (TP20) P4 fasteners: smooth nails E28/50 number of specimens: mono: 3x2 cyclic: 3x6 dynamic: 3x3 near fault: 3x3 1n 2n 3n Fig. For the purpose of connection testing the coniferous 1st class timber (JUS) is estimated as S-P-F C30 (EN).e. 136 .C5. All specimens were tested in normal weather conditions with adequate moisture control. Data on monotonic and cyclic response of joints are basic and valuable source of information needed for the development and calibration of inelastic computational models of shear walls and TFB Additional experimental data. prEN 3124). Experimentally obtained load-slip loops and post failure visual inspection of tested joints are the main source of information regarding joint response under different loadings.7 fractil 0. The correlation of materials properties between JUS and EN codes was proved by tests EN 408.D. could give important information needed for its correction and improvement.1 21.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Fig. i. The smooth nails E28/50 (DIN 1151) were applied at a distance of 10 cm.6 11. ISO 3131 (Table 1). Fig. The length of specimens depended on the number of applied fasteners in a row (1.0 strength class exp pop C40 C30 ρ05 (kg/m3) (kg/m3) 470 370 C30 330 ρ 445 360 ρ05 (kg/m3) 15. under "dynamic" or "near fault" loading protocols.5 11. The particleboard sheathing is estimated without testing.5 8. 3 nails in a row.8 29. The applied smooth nails E28/50 of Bulgarian production were certified according DIN 1151.1: Connection test specimens with 1. sample sample / density 1 2 3 3 435 ρ (kg/m ) aver. The properties were tested on 3 samples with 15 specimens in each in order to estimate the necessary data about timber strength class for ongoing tests and in general for adequate timber population. Portugal November. Testing program and Test Setup 2.2 40. by codes' nominal characteristics.0 [N/mm2] strength class acc.05 exp pop 40. i. Table 1: Mechanical properties of tested JUS 1st class coniferous timber.1 Material testing All applied materials in test specimens are available on the regional market and are regularly used by wooden house manufacturers in the region. mechanical properties bending strength fm [N/mm2] bending elastic modulus Em [kN/mm2] compression strength || to grains fc.8 C40 C27 ρsr (kg/m3) = 450 ρ05 (kg/m3) = 420 C40 C30 2.63 C40 C30 ρ (kg/m3) ρ05 (kg/m3) C27 C24 38.2 Joints specimen configuration Specimens for connection testing were made from 40/150mm C30 timber frame and P4 CEN double sided particleboards (d=12mm). 2. density in exp. Applied 3-horizontal layers particleboard PB TP20 (JUS.1. value exp pop 54. 2006 different loading protocols was done. 2.0 30.031-032) is adequate to P4 (CEN. The tests were conducted parallel to the grain. 2. 3).1 29.
as additional and comparative test.3 Testing Equipment The specimens were tested in the laboratory of the Faculty of Civil and Geodetic Engineering.5. Portugal November.1 mm/s and in the second one 0. with capacity of 100 kN. Each displacement level was repeated three times.5 0 -0. Within the first phase the loading rate was 0. Special holders with punch metal plates. Cycling load protocol consisted from two parts that divided tests in two phases. and to provide uniform distribution of acting force.5 0 100 200 300 Time [s] 400 500 Amplitude [mm] 30 1st part of loading protocol (loading rate 0. Fig. 137 1003 .4 Loading protocols Specimens were tested according to EN 26891.5 mm/s) 600 800 1000 Time [s] 1200 1400 Fig.4: Checking of moisture content in specimens.2: Detail of holder Fig. were used for gripping the wooden part of the specimen. The protocol was designed following the idea of applying significant number of cycles under the yield limit of joint and post yield cycling by following linear ascending ramp of amplitude. was used for the experimental investigation.3: Specimen fixed in with punched metal plates. Nails were hammered in wood parallel to the grain direction.2mm/s with idea to estimate the measure of max deformation capacity of the specimens and ultimate strength. Fig.2.5 mm/s. 2006 2. The Roell Amsler HA hydraulic jack press. Fig. the test continued monotonically until joint failure. After reaching the load-carrying limit of the joint. Applied loading protocols were monotonic and deformation controlled quasi static cyclic tests for ordinary and near fault ground motion. [5]. Only the upper half of the sheathing plate was fixed to timber frame element with tested fasteners. Here "dynamic" test was performed similar to cyclic. University of Ljubljana.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.3. 2. Monotonic test was applied with loading rate of 0. testing machine. The lower part of the board was glued to wooden element to provide higher shear resistance than the upper part.5 -1 -1. Fig.5 1 Amplitude [mm] 0. 1. Fig. with higher loading rate and frequency. Relative displacements are measured with inductive measuring devices positioned on both sides of test specimens. prEN12512. prEN 1380.5: Quasi-static cyclic loading protocols.1 mm/s) 20 10 0 -10 -20 400 2nd part of loading protocol (loading rate 0.
2 Mono_1. 138 . while 2n curve shows the trend of connection "set up" and lesser values.1 Monotonic test The monotonic tests were used to determine the reference displacements and ultimate deformation capacity on the basis of average values of experimentally obtained data from three samples with two specimens in each.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.74 23.79 22.3 nails_av 1 Fu. vu Ff.9mm ~20mm Loading rate was of 0. The stiffness of single hysteresis loop was defined as straight-line approximation between the positive and the negative peak of a hysteresis loop (Eq. 7 and Table 2). 1. vf Table 2: Mono . joint slip relationships and other relations derived from hysteresis loops are presented. 2006 30 25 20 15 slide [mm] 10 5 0 -5 -10 -15 0 50 100 150 200 time [s] 250 300 350 400 Near-fault loading protocol speed 0.93 13.reference values sample Fu [kN] vu [mm] Ff [kN] vf [mm] vy [mm] slip [mm] 0.01 16. * The value of yield displacement was basic for cyclic protocol accord.6 1Bn_m 2Bn_m Load / nail [kN] 0. typical load vs. Curves 1n and 3n show good coincidence near reaching ultimate load.2 0. regarding different number of nails in a row (Fig.98 14.6 0.0 0. Fig.2. prEN 12512 3. 3.35 av. 1): Ki = i i Fmax − Fmin i i δ max − δ min (1) where Fmax and Fmin are peak values of the attained load in the observed positive and negative cycle and δmax and δmin are the corresponding joint slips. Test results and comparison The objective of the performed tests was to evaluate differences in TSW frame-sheathing joints performance resulting from applying different loading protocols.43 0 0 5 10 15 20 25 30 Fig.9 0.7: Effects of nail's number in on monotonic response.2 Cyclic test Hence the cycling response of joints is the main source of information regarding structural hysteretic behaviour.5 0. which are distinguished by high-energy pulses that occur near the earthquake source.8 0. Portugal November. It was determined according to CUREe recommendations on the basis of monotonic max deformation capacity of the specimens.78 22.1 0.49 3n 0.81 23. 3.5 mm/s. vf = 22. The envelopes were calculated as average envelopes of positive and negative branch of envelope obtained by testing.99 12. 0.2 1n 1. Fig.6.45 2n 0.4 3Bn_m Bn_m 0.5 mm/s The "near fault" loading protocol is intend to model near fault ground motions.9 0.6: Near fault loading protocol. Each type of the tested joints is presented with the average envelope calculated from hysteretic responses of three samples with six specimens with the same number of nails.3 0.
4 Fu and tangent on the upper part of the envelope.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.Ductility is defined as ratio between ULS and YLS slip.7 0. Tested 3n joints exhibit higher ductility.57 0. Table 3: Parameters of average hysteresis envelope (F.3 nails in row 1n_CYC_1 2n_CYC_1 3n_CYC_1 1n_CYC_2 2n_CYC_2 3n_CYC_2 1n_CYC_3 2n_CYC_3 3n_CYC_3 3th cycl 2nd cycl 1st cycl Slip (mm) Fig.9 25.3 ULS Fu [kN] 0.5 0.δ u Ky F y. 2006 The equivalent viscous damping was calculated from hysteresis using the equation: ν eq = Ed 2π ⋅ E p (2) where Ed is dissipated energy per cycle and Ep is the available potential energy.6 0. The lines are the secant of the skeleton curve defined by points at 0.2 YLS Fy [kN] 0.60 0.8 0.8: Effects of number of nails in a row trough comparison of average envelopes of all cycles.6 Fy/Fu 0. while 2n joints have good agreement with previous testing of the same type and procedure [6].9 Fu . YLS was defined as the point of intersection between two lines.4 0.47 0.9: Definition for Yield Limit State (YLS). The definition of Yield Limit State (YLS) and Ultimate Limit State (ULS) used here are shown in Fig 8. The first one concerns major events observed from envelope curves.47 0.3 1005 139 . 3n joints show decrease of load capacity and stiffness at higher magnitudes.force.79 0. All sets of parameters are discussed regarding the influence of the number of nails in a row. Three sets of parameters obtained from the cyclic test results are chosen for detailed discussion.4 Fu and 0. Fig. The third one is related to the equivalent viscous damping calculated from hysteretic loops.36 Load. Name of specimen 1n_cyc 2n_cyc 3n_cyc Ky [kN/mm] 1.83 0.1 F u S lip .9 0.1 Fu (10 % of maximal horizontal load capacity) and 0.18 0.4 F u 0 .4 0.joint slip).94 1. The applied definitions are adopted as suitable for prEN12512 criterion for nonlinear curves.75 δu [mm] 7. The second set of parameters describes the load degradation and stiffness deterioration.9 F u In Figure 8 the average envelopes of all three cycles for each tested joint configuration are presented.first cycle. δ // // 0 . F Load / nail (kN) δy [mm] 0. which is parallel to the secant through the skeleton curve at 0.5 0.3 0.2. 0. Portugal November. δ .82 7. The basic information about the parameters of average hysteresis envelopes is summarized in Table 3.1 0 0 1 2 3 4 5 6 7 8 9 10 Cyclic testcycles' envelopes for 1.2 0.δ y 0 .5 14.48 Ductility du=δu/δy 19. F u.47 7.
Fig 12.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.12: Normalized stiffness deterioration.13. Portugal November.2 Strength degradation 1.3 1n_cyc 2n_cyc 1.7 2. Stiffness deterioration curves are very similar and have shapes of power function. Fig. Levels of 0. 3.8 ⎜ ⎟ ⎜δ ⎟ Ky ⎝ y⎠ −0. Equivalent viscous damping in tested connection was calculated from the first cycle and is in the range between 0.5 δ/δy 0 20 25 Fig.7 Strength degrad F2: F1 .9 0. The method for YLS determination used in this paper shows that YLS is in all the tested cases above the 0. Proposed equation (3) may be useful for the construction of "synthetic" load . as it was reported in [8]. Fig 11.8 of the first cycle load give information needed for the definition of YLS and ULS according to known concepts. respectively. 2 Suggested equation 1.4 1.1 3n_cyc 1n_cyc 2n_cyc 3n_cyc from first to second cycle F2 : F1 from first to third cycle F3 : F1 1 0.44x-0.76x-0.7 Slip (mm) 2 3 4 5 6 7 8 9 10 0. which is especially obvious on 3n specimens.8 1.23 and 0.9 0.8 0.7 y = 0.72 Slip (mm) Fig.4 2. 2006 1.10: Load degradation from the first to the second and third cycle for different number of nails in a row.5 1n 2n 3n where Ky and δy are stiffness and joint slip at YLS.1 1.8 ⎜ ⎜δ Ky ⎝ y 0 5 10 15 ⎞ ⎟ ⎟ ⎠ −0. F3 : F1 K (kN/mm) Stiffness deterioration K_1n K_2n K_3n y = 0.74 0. The initial elastic stiffness of the joint is relatively high.3 3 2.95 and 0.slip curve for the joints with inelastic linear behaviour from the very beginning.6 0.5 1.10. 3. but after several cycles in the range of low slip amplitudes the stiffness drops to the one that defines YLS.3 0 0 1 2 3 4 5 6 7 8 9 10 Diagrams of load degradation. 140 . Obtained power function from performed tests coincides very well with previously suggested equation in [6]: ⎛δ ⎞ K = 0.5 1 ⎛δ K = 0 .5 (3) K/Ky 3 Normalized stiffness deterioration 2. The presented diagrams can serve as aid for the definition of three linear envelope curves that is a valuable simplification for modelling inelastic behaviour of joints [7].11: Stiffness deterioration approximated with power curve. descending shape in all the cases of tested specimens and in both comparisons of test cycles. have roughly linear.95 limit.6 0 1 Fig. All stiffness degradation curves were normalized by stiffness and joint slip at YLS.2 0.
17.5 Dyn / Cyc envelopes 0.8 -15 -10 -5 0 5 10 15 20 25 30 Fig.13: Comp.6 displacement (mm) -0.3 0. Post cycle behaviour of dynamically tested joints was short.5 Failure modes and visual inspection Fig. Ultimate strength is nearly equal in all tested connections for both types of loading.e. with some tear out.2 0 -0.8 1 3n_NF : 3n_mono 0.6 0. i. of near fault and mono for 3n. Generally.4 load / nail (kN) 0. 15.5 2n_NF : 2n_mono 0.2 1 Specimens' response on near fault loading protocol is shown at Fig.16: Comp. Backbone curves have similar trend with most approximate shape in 3n joints.2 -0. 1007 141 . (b) pull trough.1 slip (mm) 0 0 1 2 3 4 5 6 7 8 9 10 Fig. mostly half of nails failed due to fatigue after forming plastic hinge inside the wooden element in the depth approximately equal to the thickness of the sheathing plate. Nails failed with pull-through of nail head while the nails exhibit plastic deformations. 1. In general.5 0.4 CyC_av Din_av 0. with the significant difference in 2n connections.8 load / nail (kN) displacement (mm) -1 -15 -10 -5 0 5 10 15 20 25 30 displacement (mm) -1 -15 -10 -5 0 5 10 15 20 25 30 Fig. 13 it can be seen that similar ultimate strengths are achieved in both test. 3. of near fault and mono for 1n. cyclic and near fault testing. load / nail (kN) 0. but stiffness deterioration between cycles is smaller in dynamic test. For monotonic. Initial stiffness is greater for monotonic load.7 0. During the dynamic protocol. An example for each of these failure modes and shapes of failed nails from observed testing is given in Fig.6 0. Monotonic testing protocol produces higher deformation capacity than near fault protocol in all the tested joints.8 0. 3. Initial stiffness is slightly greater in cyclic testing.14: Comp.2 0 -0. Detailed discussion will be omitted.5 -0. four different failure modes could be observed in the case of different testing: (a) pull out.9 0.3 0.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.4 0. 15: Comp. of dyn and cyc envelopes.2 -0.4 0 -0. but from Fig. Portugal November.6 Load / nail (kN) 0. fastener failure was predominantly pullout and pull through. 14.2 0. and (d) fracture [9].8 Dynamic test Results from dynamic testing are directly comparable with the cycling ones because of performed similar protocol with higher loading rate. Characteristic holes in timber and damage on sheathing panel are also presented in the same figure.6 -0. all failures occurred after the last cycle defined in the load protocol. 1. (c) tear out.4 1. 16 in comparison to monotonic tests for each type of connection. failure occurs very soon because of nail fracture in most cases.4 -0.2 1 Near fault test 1n_NF : 1n_mono 0. 2006 3. of near fault and mono for 2n.
pp. Skopje earthquake: 40 years of Europ. Touliatos P. Zarnic R. Stuttgart. Uang CM. Krawinkler H. Medina R. Parisi F. Centrum Hout.. Portugal November. "Development of Testing Protocols for Wood Frame Structures". McGraw-Hill. "Effects of Loading Protocol on Cyclic Response of Woodframe Shearwalls". 1995. USA. 774 pp. Holzbauwerke Grundlagen Entwicklungen erganzungen nach EC5. Their financial support is gratefully acknowledged.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 2003.: SE 40EEE. Germany. Report of CUREe Caltech Wood frame project. Wood engineering and construction handbook-Diaphragms & Shear Walls. 5. pp.. [1] [2] [3] [4] [5] [6] [7] [8] [9] References Ceccotti.. Colorado. 142 ... 2004. 2003. some general conclusions about the behaviour of the tested joint configurations with different number of nails in a row can be derived. "Blind prediction of seismic response of timber-frame house". Macedonia.A. Inc. Journal of structural engineering ASCE. Kocetov Misulic T. Proceedings of CIB-W18/36-7-5. . Faherty K...16: Nail shapes and mode of failure. 1995. . Science and Sport of the Republic of Slovenia and the Republic of Macedonia supported cooperation with experts from collaborating countries. Earthquake Eng.. 6 p. Acknowledgment The testing program was carried out in Slovenia within bilateral project with Republic of Macedonia (SI-MK 3/2002-2003).. Trough analysis of obtained results it can be seen that realistic choice for test specimen is 3n model.. Williamson T.. STEP2 to EC5. Wakashima Y. Conclusion Although the main purpose of the tests was to obtain the data needed for mathematical modelling of joints. Stanford university. 2002. The Netherlands. 6. Lecture 16.D10/10. 89p.. Gramatikov K. "Detailing of timber structures in seismic areas". 10 p. Ayaba A. "Comparison of Hysteresis Responses of Different Sheathing to Framing Joints". pp. "Design of timber structures in seismic regions". 2003. CA. 1384-1393. Timber Engineering: Design-Details and structural systems. USA. Dujic. 2001.. Germany.. Lecture D10. Gatto K. 12 p Ceccotti A. USA. 4. 16/1-16/19. pp. For all applied loading protocols 3n specimens had a good coincidence in shape of backbone curve with common 1n specimens.. B... Two nails joints exhibited quite good numerical coincidence. STEP3. but with different curve shape. Ministries of Education. Sonoda S. 551-558." Evaluation of seismic performance of nailed joints by pseudodynamic test".: "Timber shear walls . Proceedings of International RILEM Symposium: Joints in Timber Structures. Zarnic R.. Ibarra L.sheathing to framing joints" Datasheet on Robustness / Members &Connections for COST C12 -WG2. 2006 M CYC D NF Fig. Dujic B. . Skopje. Fachverlag Holz. 1989. D10/1.
aluminium and timber structures exposed to static and dynamic loads. The basic formulations and issues regarding implementation of this method are described herein. Portugal November. Secondly.arne. Introduction The Pseudo-Dynamic (PD) test method or the On-line Computer-Controlled Method of Dynamic Testing of Structural Systems has been developed during the last decades. members and sub assemblages to ascertain their strength and ductility. In the following. taking the effect of the numerical solution scheme into account. The error sources and their cumulative effects on the structural response are considered analytically. These errors are introduced into the numerical computations at each step of the solution process of the governing equations. The present investigation demonstrates that accurate tests with respect to the modelled structural system may be performed using a pseudo-dynamic test system. 2006 Pseudo-Dynamic Test Method. Simple formulas for error estimation are presented for random and systematic error characteristics. PhD. Summary The pseudo-dynamic test method simulates dynamic structural response by means of a combination of a numerical model and an experimental model in the laboratory. The combination of experiments and numerical modelling has been of particular interest in his work. 1. Methods for error detection. an introduction to the PD test method is given. In addition some examples of verification tests of the PD test method are briefly described.1 Quasi-Static Tests This is the most widely used method for determining the behaviour of joints. removal and compensation are also considered. The structural stiffness is continuously measured on the deformed experimental model. while inertia and damping effects are represented in the numerical model during execution of the test.no The author has worked with structural engineering and structural behaviour of steel. of Science and Technology 7491 Trondheim Norway kjell. The paper presents some findings of an analytical and experimental study regarding the propagation of experimental errors during a pseudo-dynamic test. The loads or displacements are 143 . Experiments and Accuracy Kjell Arne MALO MSc. Excellent agreement has been found between the observed and estimated error amplitudes. Methods for Experimental Dynamic Simulations 2. Usually a priori cyclic loading or displacement patterns are applied to the specimen to determine these quantities. Norwegian Univ. 2.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. The effects of possible experimental errors on the test results may be estimated quite accurately. Professor NTNU. first a very brief description is given of some available experimental methods for simulations of the behaviour of structures subjected to time dependent loads.malo@ntnu. It has been shown that the PD test method gives very accurate and reliable results provided that certain conditions are fulfilled regarding the experimental set-up and the instruments used to control the simulations.
Both force and displacement control may be used in the elastic range. there may be impossible to perform seismic testing of structures by other means than a shaking-table when the real time is an important parameter. the test structure is installed on the table with its mass attached.2 Shaking-Table Tests A number of earthquake simulators or shaking-tables of various sizes are available throughout the world. Typically. 2006 applied on the specimen by means of one or more hydraulic actuators which may be controlled either by force or displacement command signals. the 144 . From quasi-static tests. Portugal November. In a PD test the dynamic equilibrium of a MDOF discretized structural system is enforced only at discrete time intervals. the ductile behaviour and some energy dissipating properties of the tested specimen are found.[2] [3]. the shaking table capacity is about the same as the weight of the table. R n . 1 Conceptual flow of the Pseudo-Dynamic test method. velocity. and Fn are the acceleration. 2. Using a numerical time integration method. Since the models usually have to be much smaller than the prototypes. The shaking-table test is considered to be very reliable though there might be some problems with table . but the effect from the specimen's nonlinear behaviour on the overall response is not obtained since there is no interaction between the specimen response and the predetermined loading sequence. It is also evident that many local effects in for example connections and manufactured products as well as workmanships cannot be scaled in a sufficient manner. For such a procedure the basic equation of motion for the step-by-step solution reads [4]: Ma n + Cv n + R n = Fn Fig. while the restoring forces R n are directly measured from the specimen. but for inelastic behaviour of the specimen displacement control is commonly used since force control may lead to instability.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. In a typical shaking-table test. v n . Nevertheless.3 Pseudo-Dynamic Tests The Pseudo-Dynamic (PD) test method or the On-line Computer-Controlled Method of Dynamic Testing of Structural Systems is a relatively new method for dynamic tests [1]. Japan has some of the largest earthquake simulators which are capable of testing structures with 1000 tons mass at strong earthquake levels. 2. (1) in which the M and C are the mass and damping matrices. it is necessary to scale both the masses and the accelerogram in order to preserve dynamic similitude.specimen interactions. An earthquake accelerogram is then used as input signals to the hydraulic actuators which are responsible for the table motion The test is performed in real time which makes the observations of the structural behaviour difficult and expensive. a n . restoring force and external force vectors respectively at time nΔt where Δt is the time step used. for example the central difference method. and their usefulness is therefore limited since only complete structures can be tested realistically and the expenses of these tests are high. The mass matrix M and the viscous damping matrix C are analytically (numerically) prescribed in a computer model. The specified loading pattern may have been determined by means of a computer simulation or simply being specified as a number of load reversals at each level in a sequence of varying displacements levels.
is divided into equal time intervals Δt . Δt . structural restoring force vector. Eq. although it is basically a continuum.prior to a test some of the damping properties of the system need to be numerically prescribed. This process is illustrated in Fig. (2) can conveniently be solved by means of a direct step-by-step integration method to obtain the displacement response to any arbitrary external forcing function. d(t ) : acceleration. is that the dynamic behaviour of a structure can be accurately represented by a discreteparameter model that has only a finite number of degrees-of-freedom.} a(t ).the loads are applied on the specimen as a sequence of slow loading and pausing which implies that possible velocity effects (strain rate) on the hysteretic behaviour of the structure cannot be modelled. In step-by-step integration. Portugal November. in matrix form. By considering the equilibrium equation. • Time discretization . Also the Coulomb damping inherent in the system will be correctly accounted for. 1. Eq. • Assumed damping characteristics of the system .the governing equations of motions are treated as difference equations and solved discretely. as in most computer-based dynamic analysis procedures. 2006 velocities and accelerations can be approximated by the displacement vectors at consecutive time steps. is Ma(t ) + Cv (t ) + R (t ) = F(t ) (2) where M. F(t ) : external force applied to the system. The Basic Pseudo-Dynamic Test Method 3. velocity. the duration t for which the structural response is to be evaluated. 3.the test structure. since the external loading is prescribed and the restoring forces from the previous steps are experimentally measured.1 Basic Assumptions The formulation of the PD test method for structural systems is described in details in for instance [1] [4] [5][6]. and assuming that the solution in each step is a function of those in the previous steps. and displacements vectors. is assumed as a spring-mass discrete system. at time equal 0. The hysteretic damping is automatically incorporated in the measured restoring forces. This gives an explicit expression for the next displacement step which depends on information from previous steps only. One of the most general numerical integration methods in structural dynamics is the Newmark Algorithm [7]: Ma n+1 + Cv n+1 + R n+1 = Fn+1 (3) 1003 145 . • Quasi-static loading . The PD test method includes various approximations in its algorithm and among those are: • Space discretization .2 Pseudo-Dynamic Test Algorithm The basic assumption in the PD test method. The PD test method may briefly be characterized as a numerical simulation by means of a computer where the stiffness relationships are directly measured from the deformed specimen. v (t ). (2). C : mass and damping matrix. we can obtain an approximate solution of the equation of motion. 3. The equations of motion for such an idealized multiple-degree-of-freedom (MDOF) model can be expressed in terms of a family of second-order ordinary differential equations which. 2Δt . R (t ) . ⋅⋅⋅ nΔt .International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.
In such cases it is necessary to perform two transformations. Portugal November. 4. The basic steps of a PD test are shown in Fig. (4) and (5). 4. respectively. first from the computed displacements d to the transducer displacements Y d = Td Y 146 or Y = Td−1d (9) . Further considerations about modelling of inertia are made in [4] [5] [8] as well as in most textbooks [9] [10]. either equivalent viscous damping or addition of numerical damping forces to the restoring force vector R n+1 may be used. (3) can be solved step-by-step: d n+1 = d n + Δt v n + Δt −1 2 ⎡⎛ 1 ⎤ ⎞ d n+1 = d n + Δt v n + Δt 2 ⎢⎜ − β ⎟ a n + β a n+1 ⎥ ⎠ ⎣⎝ 2 ⎦ (4) 2 an (6) (7) a n+1 = ⎡ M + Δt C ⎤ ⎡Fn+1 − R n+1 − Cv n − Δt Ca n ⎤ 2 ⎦ ⎣ 2 ⎣ ⎦ v n+1 = v n + Δt ⎡ (8) ⎣(1 − γ ) a n + γ a n+1 ⎤ ⎦ This explicit integration method is conditionally stable. 2. The weight factors β and γ are selected by the user to achieve desirable stability and accuracy properties. By insertion of β = 0 and γ = 1 2 into Eqs. One of the main features of the PD test method is that the whole set of numerical modelling techniques is made available for the structural testing. structures where most of the mass is concentrated at few locations are best suited for testing by the PD test method. The dynamic characteristics of a discrete-parameter system may differ from those of the prototype structure since the higher modes will be truncated and the lower modes may be distorted.2 Geometric Transformations It is sometimes convenient that the numerical DOFs do not coincide with the locations of the experimental displacement transducers and the load cells used to measure the restoring forces in the deformed structure. 2006 v n+1 = v n + Δt ⎡ (5) ⎣(1 − γ ) a n + γ a n+1 ⎤ ⎦ in which a n+1 . Structures which have their masses concentrated at few locations may easily and very accurately be modelled as a discrete-parameter model. For instance. Structural Idealizations 4. various types of loading including ground excitations and wave loading can be used [5]. where ωm is highest angular eigenfrequency in the structural system. ωm Δt < 2 . Proper models of such structures demand use of a large number of DOFs which increase the complexity of the experimental set-up.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. In order to obtain desirable damping. Therefore. Consequently. an explicit integration method is obtained. Here. the degrees-of-freedom (DOFs) need not to be selected in order to represent the stiffness relationships of the test structure exactly since they are directly measured from the deformed specimen in terms of restoring forces at the specified DOFs. velocity. By selecting β = 0 . at time equal to ( n + 1) Δt . while systems which have an almost uniform mass distribution are more troublesome.1 Discrete Parameter Model A structure has to be idealized as a discrete parameter system in order to be tested by the PD test method. v n+1 . As a consequence. Eq. and displacement vectors. and d n+1 are the acceleration. Unlike other numerical simulations. the imposition of rotations and the measurement of moments are cases where there is not a one-to-one relationship between the numerical DOFs and the hydraulic actuators [6]. an explicit integration method is defined as a method where the displacement solution in each step is a function of previous steps only. only dynamic properties such as the inertial distribution need to be considered besides the external forces. for instance by means of the finite element method.
the tested structure may be subdivided into two parts in which one substructure is physically tested while the rest of the structure may be modelled numerically. R n is the nodal forces originating in the experimental model. More information about this sub structuring technique may be found in [11]. while nonlinear deformations will occur in the 1005 147 . The numerical model may of course be nonlinear. while R n = Kn d n is the contribution from the numerically modelled part. Some types of structures need not to be tested with their full design weight included. Such angular changes will always lead to errors in the measured forces. the total restoring forces may be found from: E N E N Rn = Rn + Rn = Rn +K n dn (11) E N N Here. some offshore jackets are designed such that the main legs which carry the dead loads behave linearly elastic during severe earthquake excitations. Portugal November. For large displacements.3 Restoring Forces The measured forces may directly correspond to the nodal restoring forces in the model or be computed as the sum of experimental and numerical contributions. When the compatibility conditions between the two parts are taken care of. 4. 2 Basic experimental scheme for the Pseudo Dynamic test method. For instance. 2006 and secondly from the measured forces in the load cells P to the nodal restoring forces R : R = TR P (10) In general Td and TR are different matrices. it may be necessary to let Td and TR be functions of the displacements in order to correct for angular changes in the actuators. [5] while an example on application of this concept is given in [6].International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. while the errors in the imposed displacements may be minimized by use of long and suitably positioned displacement tranducers such that the angular changes become insignificant. For instance. Fig.
The advantage with equivalent viscous damping is that the stability limit of the explicit integration algorithm is unaffected. Error propagation in the higher modes may cause serious problems in PD simulations due to their higher error amplification factors. The fact that the restoring forces are proportional with the square of the frequencies. Hilber et al.n+1 . 1 . cracking etc. K n is a geometric stiffness matrix and R n is the modified restoring force vector. but the disadvantage is that this kind of numerical damping will reduce the stability limit of the integration algorithm. the material damping due to hysteresis. it is still desirable to take the overturning effect of the top weight into account and this may be done by a modification of the restoring force vector in terms of the current displacement pattern: G R* (12) n = R n +K n d n G * Here. However. 4.4 Damping and Rate Effects The damping forces FD in a structural system may be due to various physical effects like friction (Coulomb damping) and viscous effects. Portugal November.4. as well as in most numerical simulations.2 Numerical Damping The explicit version of the Newmark integration algorithm does not contain numerical damping. Structures exposed to earthquakes are sometimes also equipped with discrete dampers to absorb the earthquake imposed energy or to transform the frequency content of the energy input to frequencies outside the vulnerable frequency band. 4. 2006 bracings. will be inherent in the restoring force vector R n+1 and is hence automatically incorporated. (3) we may replace the viscous term Cv n+1 by FD . Examples of applications of this concept may be found in [1][4]. In a PD tests with several DOFs. while larger damping may be imposed on the higher modes. In order to use equivalent viscous damping as en error compensation method. knowledge of the stiffness matrix is not necessary. If their properties are known they may be modelled numerically and added to FD . However. The damping modelled by FD is velocity dependent. It is also possible to account for stress or strain rate effects in FD .n+1 = α ( R n+1 − R n ) + ρ Δt 2 M ( d n+1 − d n ) (15) The parameters α and ρ controls the amount of numerical damping which might be approximately proportional to stiffness and mass.4. plasticity. the familiar Rayleigh damping is obtained. 4. In Eq. where FD also may contain contributions from other damping mechanisms. make them sensitive to error growth in the higher modes. Based on this concept Shing and Mahin [8] developed an explicit version of the Newmark algorithm for PD testing which reads: FD . a small amount of damping should be applied in order to avoid growth of spurious higher frequency response. it is necessary to apply frequency dependent damping. The damping matrix may then be taken as a Caughey series [12] with: FD . [13].n+1 = Cv n+1 where: C= ∑ α k M ( M -1K ) k k (13) Using only the two terms k = 0.1 Equivalent Viscous Damping For MDOF systems frequency dependent damping is preferable as it allows a small positive. Here.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. while the disadvantage is that it requires knowledge of the stiffness matrix K. negative or zero damping to be prescribed for the fundamental mode. 148 . showed that introducing an approximation to the velocities n+1 = v d n+1 − d n Δt (14) numerical damping could be obtained.
However. no iterative techniques on the specimen is possible. 3. Due to the numerous steps involved in the quasi-static simulation the cumulative effect of feedback errors may become important. but the test will still be executed with an expanded time scale. using the measured information R n+1 . 5. experimental feedback occurs in the step-by-step numerical integration. The restoring forces R n+1 associated with the new deformed state are measured by the load cells. 4. Experimental Error Effects 5. various expansions or variants of the PD test method have been developed [14] [15]. 2: 1. For a the inaccurate measurements of the restoring forces denoted e rm n linearly elastic system the measured restoring forces may be expressed as [12] 1007 149 . In PD tests.4. In static or quasi-static tests these experimental results are commonly acceptable if the errors at each step of the testing procedure are within specified limits. Secondly. Accelerations a n+1 and velocities v n+1 are computed by means of Eqs. In a pure numerical simulation it is possible to use implicit time integration algorithms in combinations with equilibrium iterations to compute the next time step. The main advantage in a basic PD test using loading and pausing is more control with the test conditions and that conventional test equipment present in most laboratories might be used without modifications. 2.1 Error Characteristics No matter how carefully an experiment is designed and executed. To incorporate some of the rate effects. effects of inertia both from the test specimen as well as the test equipment would be present in the measurement of the restoring forces and might lead to overestimation of inertia effects. Possible rate effects in the restoring forces are consequently not present in the measurement. These rate effects may be accounted for numerically if their characteristics are known. and thus the test results may diverge from the correct ones.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. (7) and (8) respectively.3 Rate and time scale effects The PD test method is basically a quasi-static test which is performed in a stepwise manner with loading and pausing leading to a prolonged time scale relative to the real time scale. These displacements are imposed on the laboratory specimen by means of the hydraulic actuators. It has been sought to overcome this problem by using extrapolation and interpolation techniques. The measurements of the restoring forces are performed during the pausing periods to assure that the correct displacement patterns are imposed on the specimen (after a user specified time delay). this test strategy is not consistent with an explicit time integration method since it is necessary to know the restoring forces from the previous steps in order to compute the displacements to be imposed in the next step. However in a physical test. It is obvious that the experimental errors introduced into the equations of motion originate from the imposition of d n+1 and the measurement of R n+1 and enter the equations of motions as errors in R n+1 . 2006 4. or some kind of predictor – corrector schemes in the numerical integration algorithm. giving continuously performed tests without stops. but still at an expanded time scale. due to the fact that nonlinear structural response is path-dependent. The displacements d n+1 for time step n + 1 are computed by means of Eq. Steps 2 and 3 are the only ones that distinguish the PD test method from a traditional numerical simulation. however. Due to improper displacement control at step n displacement errors e dc n are introduced. (6). Typically the time scale might be expanded with a factor in the range 100 – 1000. The basic steps of a PD test are outlined in Fig. It seems to be necessary to integrate the actuators electro-hydraulic servo-controllers with the numerical integration scheme of the pseudo-dynamic test in order to perform a continuous PD test. The main idea has been to avoid the pausing periods. The advantage in a continuous PD test is that some rate effects will be incorporated in the measured restoring forces. Portugal November. give rise to the errors. experimental errors are present in the test results. and error propagation may hence take place. If a PD test is executed at an elevated speed close to the real time scale.
Hence the experimental errors are of less significance for inelastic PD tests than for elastic tests. Stiffness errors are most probably due to inaccuracies of the support conditions or miscalibration of the relationship between voltage and the physical quantity measured. such errors may to some extent be compensated for by adjusting the numerically prescribed mass matrix such that the correct frequencies are obtained. These errors are usually frequency independent. Portugal November. they will contain errors both from the imposition and the measurement dc dm ed n = dn − dn = en + en (19) The characteristics of experimental errors have been investigated [12]. These errors may be frequency dependent due to interactions between several actuators. friction in the mechanical parts of the test setup. 2006 ˆ = K ( d + e dc ) + e rm R n n n n (16) ˆ is the measured where d n is the numerically computed displacement vector for step n .2 Error Quantification Mathematical tools for error estimation in PD test results have been derived in [16].2. and R n restoring force vector with feedback errors included.e. However. energy changing effects. [17] and [6] and it has been shown that the introduction of frequency dependent damping dramatically reduces the experimental error propagation. There are two main causes of erroneous energy changes in the response: • Coulomb damping i. • Inaccurate displacement control due to the imperfections of the actuator control loop and/or improperly set control parameters. and may cause erroneous structural eigenfrequencies and rapid growth of spurious higher mode effects. as well as the cumulative error amplitude developed. in cases where the errors are inphase or 180 degrees out-of-phase with the velocity response. All these errors will probably co-exist in a particular PD test. Errors in the energy of the system are due to experimental errors which are in-phase with the ( ± ) velocities. 5. However. This force feedback error may for linearly elastic systems be handled as an equivalent displacement error −1 dc rm e rd n = K ( Ke n + e n ) (18) In the case that the imposed displacements are measured for control purposes or for error estimation.1 Cumulative Random Errors The cumulative displacement errors due to random errors may be calculated by [6] 150 . For nonlinear systems a displacement drift error may occur. the relative influence of the cumulative errors on the structural response are approximately inversely proportional to the ductility level. For the latter case the errors in linear systems cause either erroneous structural stiffness resulting in distortions of the fundamental periods or. The errors in the restoring forces caused by experimental errors in step n are hence [12] dc rm ˆ ˆ er n = R n -R n = R n − Kd n = Ke n + e n (17) where R n is the restoring force vector developed by d n . since the absolute values of the input errors are not sensitive to the force and displacement levels. These errors result in hysteresis in the force-deformation relationship. 5.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. [6] and it has been found convenient to distinguish between random errors and errors which are systematic with respect to the structural response. The magnitude of the erroneous displacement drift depends on the relative phase characteristics between the error response and the response due to external forces.
(21) and 1 es( m ) = ωmt 2 .ξm > 0 er ( m ) ( t ) = σ e ωm Δt ⋅ 2ωmt er ( m ) ( t ) = σ e 2 . σ e is the errors standard deviations.ξ m > 0 (30) 1009 151 .ξ m = 0 (29) es( m ) = 1 1 − exp −ξmωmt 2ξ m ( ) . 2006 en = ∑{φ } m =1 m M 2 F( 2 e2 t m) r( m) ( ) (20) where: F( m ) = 2 1 − ξm − ξ Ω −ξ Ω exp ( m m m m ) 2 2 1 − ξm − Ωm 4 (21) (22) .International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.ξm = 0 m m ωm Δt 2 ⋅ (1 − exp −2ξ ω t − ξ m ) 2 ξ m (1 − ξ m ) (23) and {φ m } are eigenvectors where each term is squared and computed from normalized eigenvectors such that φT m ⋅ φ m = 1 . Portugal November.2 Cumulative Systematic Errors The cumulative displacement errors due to systematic errors in a MDOF structural system may be estimated by [6] en = ∑ B m A m F( m )es( m ) m=1 M (28) where F( m ) is given by Eq. In the above expressions ξ m is the modal damping ratio calculated by ξm = and Cm φT mCφ m = 2ωm M m 2ωmφT m Mφ m (24) Ω m = ω m Δt = Km φT m Kφ m Δt = Δt T Mm φ m Mφ m (25) Their numerical counterparts read ⎛ 1 − ξ mΩ m ⎞ ln ⎜ ⎟ 1 + ξ mΩ m ⎠ ⎝ ξm = − Ωm ⎛ ⎛ Ω2 2 m ⎜ Ωm ⎜1 − ξ m − 4 ⎝ Ω m = ωm Δt = arctan ⎜ 2 Ω ⎜ 1− m ⎜ 2 ⎝ ⎞⎞ ⎟⎟ ⎠⎟ ⎟ ⎟ ⎠ (26) (27) 5.2.
Verification Examples using Simple Steel Structures 6. In Fig. have been performed [5]. The matrix B m is given by Bm = φ mφT mM T φ m Mφ m (31) 6. The purpose of these tests was to verify the accuracy of PD test system at NTNU. and no rotational DOFs were hence included in the dynamic equations of motion. 3 and Fig. On this basis a test program was chosen with the following main features: • Automatic stiffness measurement. while the elastic restraint of the test rig is represented by a rotational spring kθ . and where the model had two well separated structural eigenfrequencies in a frequency range relevant for building structures. Portugal November. a cantilever beam with two translational degrees of freedom (2DOF) and a cantilever beam with 2 translational and 1 rotational degrees of freedom (3DOF). The masses were assumed to have only translational degrees of freedom.1 Introduction Two multiple degrees of freedom systems (MDOF) have been tested at NTNU [6]. 6. rectangular hollow section (RHS) 120x120x6. Division of Structural Engineering. The total length of the cantilever was 1925 mm and the two masses were located 1925 mm (DOF1) and 1020 mm (DOF2) from the base respectively. 8. • Earthquake excitation giving linearly elastic response. Ground acceleration due to earthquake constituted the loading for the forced vibrations tests with the S69E component of the Taft 1952 earthquake used as the chosen record. as well as to gain experience with testing of MDOF structural systems. In addition two tests on SDOF systems (a simple SDOF cantilever beam and a braced frame). The masses were represented by a lumped mass matrix given by 152 . confer Fig. 2006 A m is the error amplitude vector in geometrical coordinates corresponding to vibration mode m . 3 ground accelerations is given the symbol ag . • Free vibrations using mode shapes as initial conditions. 3 Structural 2DOF model. • Earthquake excitation giving inelastic response. This lead to the choice of a simple cantilever beam with two concentrated (lumped) masses as shown in Fig. Consequently the dynamic behaviour of this structural system may be represented by a set of only two equations. The particular structural models were chosen in order to have little complexity in the experimental setup and few physical effects influencing the test results such that they could easily be compared to pure numerical simulations using simple models.3 mm.2. The geometry and the stiffness of the two structural systems were quite similar.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.2 Two Degrees of Freedom System 6.1 Structural Model The choice of the structural model for the 2DOF test was primarily determined by the desire to test a MDOF structural model where the complete model was physically located in the laboratory. The cross-section used was a hot rolled steel Fig. 3.
2. Portugal November. The resulting restoring forces developed by the specimen were measured by the load cells associated with each DOF. The base flexibility was considered to be acceptably small. The cantilever was mounted in a horizontal position and fillet welded to the vertical front-plate of a stiffened plate girder. positions of measurement transducers.2. Fig. The direct stiffness approach was utilized by imposing a displacement cycle at one DOF at a time while keeping the displacements at all other DOFs equal to zero. 2006 0 ⎤ ⎡1100 M=⎢ 1800 ⎥ ⎣ 0 ⎦ ⎡1⎤ F = −M ⎢ ⎥ ag ⎣1⎦ mass matrix [kg] (32) The external forcing function due to ground excitations for this simple system becomes external force vector (33) 6. A theoretical stiffness matrix K teo for the model may be determined using the measured material 1011 153 . 4.3 Initial Stiffness Properties The elastic stiffness properties of the structural system were determined by use of the quasi-static mode of the PD test system.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.2 Test Arrangement A schematic view of the test arrangement is given in Fig. 6. 4 Test arrangements. but showed an approximately linear behaviour though the rotational stiffness was to some extent dependent on the displacement pattern imposed. The girder did not provide a complete fixation of the cantilever. The experimental stiffness matrix K E was measured based on the forcedisplacement relationships.
The accelerogram is depicted in Fig. By solving the general eigenvalue problem for each mode m ( K − ω m M ) φm = 0 (34) it was found that K E and K teo gave about the same eigenfrequencies for the lowest mode. S69E component with peak acceleration scaled to 0. Portugal November.07 g as ground excitation.S69E component. where also the eigenfrequencies for the test structure are indicated. b) Fourier spectrum for the first 20 seconds 6. Due to the irregular nature of the Fourier spectrum small changes in the eigenfrequencies of the 154 . A Fourier spectrum for the first 20 seconds is shown in Fig. 2006 and cross-sectional properties. 5 a).4 Ground Excitations giving Elastic Response A low level elastic PD-test denoted 2PEQ-EL was performed using the Taft 1952 accelerogram. 5 The Taft 1952 accelerogram -. a) Acceleration time history. This indicates that the measurement of the stiffness terms corresponding to the second mode of vibration are more sensitive to both experimental support conditions and displacement control errors due to the small displacement levels but large restoring forces involved.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.2. Fig. while they differed about 10 % in the highest mode. 5 b).
The standard deviations due to random errors σ e were 0.14 ⎤ en = σ e ⎢ ⋅ 23.e.11⎦ ⎣ 0. 2006 structure may cause significant changes in the response.0 and ξ 2 = 0.29 ⎤ ⎡5.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.14 ⎦ [mm] (35) The experimental errors were practically constant in magnitude and did not depend on the displacement amplitudes. The 2PEQ-EL test was also simulated numerically with Δt = 0. while the viscous damping matrix C was taken equal to zero.7 + ⎢ ⋅ 39.005 s. (19) did not show any systematic contribution [6].5 Experimental Errors The experimental errors e d n Eq.2 = ⎢ ⎥ ⋅ σ e = ⎢ ⎥ ⎥ ⎥ ⎣ 0. The numerical simulated response showed excellent agreement without any visible differences to that obtained from the 2PEQ-EL test in the linearly elastic case. i. Their relative influence is therefore greatest for tests with low level excitations.89 ⎤ ⎡ 0. Fig. 6 depicts the time history for the displacements of the 2PEQ-EL test. The numerical damping was specified to ξ 1 = 0.005 s.5⎦ ⎣ 0. Experimental errors at this level have no significance for the evaluation of the experimental results. and indicates that the 2DOF system was oscillating primarily in its lowest mode. 6. 6 Displacement time histories for elastic pseudo-dynamic test 2PEQ-EL.71⎦ ⎣5.025 mm [6] and the cumulative displacement errors for the 2PEQ-EL test were obtained from Eq.6⎤ ⎡ 0.2. Fig.01 . Portugal November. The PD test was completed in less than one hour. tests where only small parts of the test equipment range are used. The first 20 seconds of the response history were simulated using 4000 time steps using time increment Δt = 0. (20) as ⎡0. 1013 155 .
65g. For comparison it should be noted that the discrepancy between the PD test response and the numerical simulation exceeded 20 mm.7 s. 6.2. This discrepancy increased with time as may be observed from Fig. many uncertainties still remain. as shown in Fig. a large displacement offset is present in both time series. A corresponding nonlinear numerical simulation using a special purpose program [18] is denoted 2NEQ-PL.1 Structural 3DOF Model The dynamic structural model is shown in Fig.6 s. The objective of this test was to couple an experimental model with a numerical model (substructure technique) and to compare the response obtained from the mixed experimental and numerical test model with the complete 2DOF experimental model described earlier in this paper. the material hardening behaviour cannot be revealed without cyclic material testing and the residual stresses from welding is more or less unknown.6 Ground Excitations giving Inelastic Response The inelastic test 2PEQ-PL was carried out using the Taft 1952 accelerogram scaled to 0. a displacement discrepancy occurred between the test and the numerical response. For comparison the displacement time histories of d1 are shown in Fig. and for this time interval the PD test and the numerical simulation showed almost identical results. 7 Displacement time histories for d1 .International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. the mass distribution and the measured cross-sectional properties.one purely numerical and one experimental. The offset is largest for the numerical response. 9. The specimen behaved elastically until about t = 3.3. although the simulated response is somewhat smaller than that of the test. 156 .3 Three Degrees of Freedom System The 3DOF test differed from the 2DOF test only with respect to minor differences in the rig systems. 8. However. After a displacement peak at about 6. and with the same input parameters as used for the elastic test 2PEQ-EL. This fact illustrates that. even in this clean and simple problem. the constitutive modelling is a challenging task. 6. after the first nonlinear excursion at approximately t = 3. Fig. Although coupon tests were performed on the test material. Portugal November. 7.6 s. These discrepancies were probably mainly caused by the constitutive modelling used in the numerical simulation program as it also failed to give correct shape of the hysteretics loops. The structural system was divided into two submodels . 2006 6. For instance. 7 for both test and simulation. inelastic test 2PEQ-PL and numerical simulation 2NEQ-PL The responses for the two cases show quite good agreement. It was also of major interest to find practical methods for the imposition of rotations on specimens by use of electro-hydraulic actuators in order to maintain the rotational compatibility between the numerical and experimental substructures.
020 m. The numerical sub-model consisted of a single beam element with two translational and one rotational degrees of freedom. The experimental sub-model consisted of a cantilever beam with one translational (DOF2) and one rotational (DOF3) degree of freedom. Due to the explicit algorithm used in the time integration of the equations of motion. The element stiffness matrix was determined using the nominal cross-sectional and material properties. kθ = 1. 6. The viscous damping matrix C was taken equal to zero.e.3. The following matrices were used: 0 0 ⎤ ⎡1100 ⎢ M=⎢ 0 1800 0 ⎥ ⎥ ⎢ ⎥ 0 0 100 ⎣ ⎦ ⎡1 ⎤ ⎥ F = −M ⎢ ⎢1 ⎥ a g ⎢ ⎣0⎥ ⎦ mass matrix [kg] (36) external force vector (37) The masses have units [kg] and [kg m2/rad] for translational and rotational DOFs respectively.2 Test Arrangement A schematic view of the test arrangement is given in Fig. 2006 Fig. 1015 157 . The experimental sub model consisted of the cantilever beam spanning from point A (the base) to B (at DOF2) with length L1 = 1. Although the geometry was identical to that of the 2DOF model. The restoring force vector R was obtained by adding the contributions R E measured from the experimental model and R N calculated from the numerical one: ⎡0⎤ E⎥ N R = RE + RN = ⎢ ⎢ R2 ⎥ + K d E ⎢ ⎣ R3 ⎥ ⎦ (38) where K N is the stiffness matrix for the three DOF numerically modelled beam element.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 9 Experimental (left) and numerical (right) submodels. 10. Fig. At the base the elastic restraint was identical to that of the 2DOF test i. a small rotational mass m3 was needed in this test.0 ⋅108 Nm/rad. 8 Structural 3DOF model. an additional degree of freedom d3 was necessary in this model in order to maintain the rotational continuity at the point of intersection. Explicit time integration algorithms are only conditionally stable and zero mass at one degree of freedom will lead to infinity large eigenfrequencies and unstable solutions. Portugal November.
One of the actuators (DOF2) was attached at point B ( P2 ) and the other at point D ( P1 ) with the distance L2 = 0.020 m from the base. The Y1 transducer was attached to a rod which was welded to the specimen at a distance L1 = 1.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.807 m between point B and point D. Fig. 2006 In order to impose both translation d 2 and rotation d3 at point B two actuators were used.942 m between them. Load cells were mounted on both actuators and the restoring forces developed from the experimental sub-model were calculated from the measured forces P 1 and P 2 . The Y1 transducer was used to control the displacements of the actuator ( P1 ) while the Y2 transducer was used to control actuator ( P2 ) . The restoring force R2E and the moment R3E may then be computed as follows: ⎡ R2E ⎤ ⎡1 ⎢ E ⎥ = TR P = ⎢ ⎣ L2 ⎣ R3 ⎦ 1⎤ ⎡ P 1⎤ ⎥ ⎢ 0 ⎦ ⎣ P2 ⎥ ⎦ (39) The displacement transducers Y1 and Y2 were located with the distance L3 = 0. The actuators were both attached to the specimen with clevises in the same manner as used at DOF2 in the 2DOF test. Portugal November. 10 Schematic view of the 3DOF test arrangement with the positions of the measurement transducers indicated. The relationship between the experimental displacements Y1 and Y2 and the degrees of freedom of the experimental sub-model may be expressed by ⎡ 0 ⎡d2 ⎤ ⎢ = T Y = d ⎢ ⎥ ⎢ 1L ⎣ d3 ⎦ ⎣ 3 ⎤ ⎡Y ⎤ ⎥ 1⎥ − 1 ⎥⎢ L3 ⎦ ⎣Y2 ⎦ 1 (40) 158 .
6. 11 between the 3DOF test specimen and the 2DOF specimen (described in the previous section) subjected to the Taft accelerogram scaled to 0. simultaneous use of rotational and translational degrees of freedom is demonstrated to be feasible in pseudo-dynamic tests. 7. a static condensation was carried out on R3 = 0 . mode 16. Portugal November.894 1 -1.3. Table 1 Comparison of modal properties obtained from the 2DOF test where the complete model was experimental.316 1 -1. has a great potential in studies of structural behaviour of both components and complete structural systems. 12. mode Eigenfrequencies [Hz] Eigenvectors φ m 3.3 System Stiffness Matrix By adding the stiffness matrix K E for the experimental submodel to the corresponding terms in the stiffness matrix K N for the numerical model the system stiffness matrix K was obtained.5564 From this study it can be concluded that the stiffness of the hybrid model showed good correlation with that obtained from the complete 2DOF experimental model. The correlation between the displacement time histories for d1 in the 3PEQ-PL and the 2PEQ-PL tests are in general good. It has been demonstrated that the PD test method is very reliable and give accurate test results provided that certain conditions are fulfilled regarding the experimental techniques. it is shown that almost identical amounts of hysteretic energy were dissipated in the two tests. the eigenvalues and eigenvectors are used as basis of comparison. Conclusions The paper adresses a quasi-statical method for testing structural systems subjected to dynamic loading by means of computer-controlled hydraulic actuators.Again. As it may be observed from the plot the 3PEQ-PL test showed smaller maximum displacement amplitudes. and from the coupled experimental and numerical 3DOF model.65g peak acceleration.4 Ground Excitations giving Inelastic Response A comparison is given in Fig. mode 16. which combines the versatility and convenience of numerical techniques with the realism of structural tests. This yielded a 2x2 stiffnes matrix denoted K 2 .5666 1. Referring to Fig.3. 1017 159 . 2006 6. The time histories from the 2DOF and the 3DOF tests were slightly offset from each other. In order to compare this stiffness matrix K with that determined from the 2DOF test.3528 2. The use of the PD test method should hence be extended and developed further to a wider range of structural systems.129 1 0. see Table 1. where they are compared to the results obtained from the pure 2DOF experimental test. Stiffness matrix from 2DOF test Stiffness matrix from 3DOF test (condensed) 1. The restoring forces showed no significant difference.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. It is hence concluded that the difference between the two tests is of the same order as expected between to different experimental tests. The pseudo-dynamical test method. Furthermore. mode 3.174 1 0.3597 2.
Fig. 11 Displacement time histories displacements at DOF1 for 3PEQ-PL and 2PEQ-PL.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 12 Energy dissipation histories for 2PEQ-PL and 3PEQ-PL tests. 2006 Fig. Portugal November. 160 .
J. and Mahin S. Vol. Inst. The Norwegian Institute of Technology. Ing.G. 1977. and Mahin S. TAPIR. S. No.: Pseudodynamic Test Method . Trans R. A (2001) 359. and Dermitzakis S..A. Phil. June 1983. Dr. Journal of Engineering Mechanics. The Royal Society 2001. February 1985.K.A. R.B. 1989. Lond. Report No. 1851-1867. Earthquake Engineering Research Center. Vol 27.Experimental Accuracy and Techniques.N. et al: Nonlinear Earthquake Response Analysis of Structures by a ComputerActuator On-Line System. Division of Structural Engineering. 7. Earthq. CA. Newmark.: Classical Normal Modes in Damped Linear Systems. and limitations of real-time online (pseudo-dynamic) testing. Resistant Struct. No.N. and Nakashima M. Dermitzakis S. 85 EM3. and Mahin. Earthquake Eng.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Magonette.K. G.A. Trondheim. Report No. Shing P. Div. 1988.: Structural Vibrations. Trans R. Vol.: Dynamics of Structures. [1] References Takanashi K.A.M.: Cumulative Experimental Errors in PDT Earthquake Engineering & Structural Dynamics.. U. Trondheim. Bergan P. Eng. 1960. Division of Structural Engineering. 1984. Mech. Bul. 1981. Inc. Shing P.R and Taylor R. and Penzien J. 2006 8. Trondheim. Earthquake Engineering Research Center. Vol..A. Earthquake Engineering Research Center.A. 13. Earthquake Engineering & Structural Dynamics. No. 1988. 4/July-August 1985. UCB/EERC-84/01. Journal of Applied Mechanics.: A Method of Computation for Structural Dynamics. University of California. N. Ing.: Computational aspects of a seismic performance test method using on-line computer control. Univ. National Conf. Ing. 113. 5. 1986. A (2001) 359. Dyn. Soc. CA. Eberg E. Shing P.: Development. UCB/EERC-85/04. Soc. 1975. Thesis.. of Tokyo. Trondheim. May 1987.: Pseudo Dynamic Testing --. A. Nakashima. Division of Structural Engineering. University of California..W. The Norwegian Institute of Technology. J.B.Numerical Algorithms and Implementation. Thesis. of Industrial Science. CA.M. Thesis. Astrup O. and Struct. Phil. Dr.: Developement of Substructuring Technics for On—Line Computer Controlled Seismic Performance Testing. University of California.B. potential. [2] [3] [4] [5] [6] [7] [8] [9] [10] [11] [12] [13] [14] [15] [16] [17] [18] 1019 161 . The Royal Society 2001. UCB/EERC-83/12. ASCE. Report No. Berkeley. Lond.: Experimental error propagation in pseudodynamic testing. Thewalt C. McGraw-Hill. of Earthquake Eng. Hilber H. M.B. The Norwegian Institute of Technology. Clough.C. 1975. July 1987. Mahin S. Berkeley. Shing P. Portugal November.: Advanced On—Line Computer Control Methods for Seismic Performance Testing. and Mollestad E.L: Improved Numerical Dissipation for Time Integration Algorithms in Structural Dynamics. Aug. Vol.: Pseudodynamic test method for seismic performance evaluation: theory and implementation. Hughes T. No. and Mahin S.: Japanese activities on On-Line Testing. 15. Dr.: Development and application of large-scale continuous pseudo-dynamic testing techniques. 8.. 1959. Larsen P.: Cyclic Plasticity Modelling in Force Space. Mahin S. Vol. Malo K.S. Berkeley.R. Caughey T. 1771-1799. Takanashi K.
cyclic loading 1. More advanced testing procedures. but concerns remain due to the velocity of the load transmission. FÜLÖP PhD. Hungary Dan DUBINA Professor. 2006 Seismic Performance of Timber Shear Walls Sheathed with OSB Panels Ludovic A. which creates a fundamentally new design scenario for the structural typologies proven to perform well in non-earthquake zones.ro Fülöp Ludovic. graduated from the Civil Engineer Faculty of the ' Politehnica'University of Timisoara in 1998. subassembly testing is common. In the Mediterranean region and Eastern Europe light-framed constructions are new. Politehnica Univ. both using steel and wood skeleton.. OSB sheathing. which are the main load bearing elements against lateral loads. are analyzed by: testing of wall details. For instance wall-panels. The results are integrated in order to identify vulnerabilities of the shear wall system and formulate improvement proposals for details and some design recommendations. As testing procedures improved quasi-static cyclic tests became the standard for earthquake loading. earthquake performance. Keywords: timber shear walls. A practical way to understand the behaviour of certain elements of a house is by testing. With this procedure the cyclic nature of the earthquake action is taken into account. One such problem is earthquake loading. due to earthquake and wind. are often tested. with lack of design codification and industry experience. Buttler LTD. VTT –Finland Ludovic. In this article the study of an existing prefabricated timber-house system is presented from the point of view of earthquake performance. which represent the primary load bearing elements of houses in case of earthquake. Because it is not easy to perform full scale test on entire structures. Designs are often based on the expertise of the structural engineer in charge but. The prefabricated shear-walls. certain aspects of the structural behaviour can not be easily predicted. Introduction Light framed house constructions. pseudodynamic and shake table tests have also been used [1][2]. He received the PhD from the same university in 2003.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.dubina@ct. Senior Researcher. because the structural scheme is complex. which is low during test compared to the real loading scenario. In the early phase of development experiments were carried out using monotonic testing procedures. of Timisoara.Fulop@vtt. Politehnica Univ. born in 1975. are becoming more popular in Europe. Summary As light timber-framed houses gain market share in Europe new design challenges emerge due to the local loading conditions in different geographic regions.upt. 163 .fi Ágnes RUFF PhD Stud. Portugal November. The aim of such tests was solely to determine the rigidity and load bearing capacity of the walls in order to supply reliable design values. of Timisoara Romania Sándor BÁLINT-MAJOR Structural Designer. Finite Element Modelling (FEM) and full scale cyclic testing. Romania dan.
but can be taken into account in combination with other panels. but also the connections between the members of the frame are important. floor or roof assemblies are prefabricated with complete finishing. The main findings of these experiments are summarised here: • • • • There is no qualitative difference of behaviour between wall panels sheathed with different plate types (i. Wood skeleton walls are usually characterised by large ductility due to the ductile behaviour of the connections.g. rigidity. and that of the sheathing itself is less ambiguous and can evaluated analytically. The building technique under investigation (Fig. 2. Supplementary gypsum board increases the stiffness and strength of the wall-panels but the ductility is usually reduced.e. connections to other building parts). 2006 Summaries of experimental research on timber-framed walls have been published by Tissel [3]. hold-down anchorages. Usually. the possibility to study the wall by means of its components is important from two points of view. 1) was used in regions with no earthquake loads and has the main advantage that entire wall. OSB. To ensure overall ductile failure of the wall a hierarchy based on the strength of the components has to be enforced in the design.e. not extremely thin or slender). Secondly. the more expensive wall panel tests can be partly replaced by test on components completed with modelling. • In conclusion. strong pinching and large ductility. Wolf [4] and Foliente [5]. The house is delivered to the site in a few pieces significantly reducing site work and construction time. Goal of the study Fig. plywood or gypsum). Firstly in order to ensure that components with significant ductility are the weakest. However. 2. the components with the most significant influence on the performance of the wall are sheathing to frame connections and hold-down details. hold-down details are usually more complex with less easily predictable behaviour. The components governing the behaviour of the wall panels are connections. ductility) of the entire wall-panel. 164 . A four step approach was used in the study consisting of: (i) small scale tests on components believed to significantly influence the behaviour of the walls. These components are also important because the behaviour of the frame element. The connections between the sheathing and the frame. highlighting the importance of the hold-down. strength. with the goal to identify potential weaknesses. 1 Typical house after finishing The focus of this study was to asses the earthquake performance of a prefabricated timber wall typology used in house construction. In such a way. Gypsum board alone is usually not recommended as load bearing sheathing. Overall. (ii) FE modelling of the wall panels. The geometrical configurations of the walls at the basis of the study are presented in Fig. the framing has little influence on the behaviour of the wall-panel.e. (iii) full scale test for some wall configurations to validate the FE models and (iv) integration of the results and evaluation of the overall performance and weaknesses.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. tests are made on isolated shear walls and uplift forces are transmitted to the anchoring alone. If the framing members are of usual dimensions (i. Other critical components of the walls are overturning restraints (e. If the framing members are weak they can trigger the failure of the wall resulting in a non-ductile behaviour which should be avoided. The disadvantage of the system is the limited detailing freedom of the connections as consequence of the high prefabrication level. Portugal November. In all cases the hysteretic loops have similar shape characterised by non-linearity. it is useful to the design process to find relationship between the characteristics of the components and the overall properties (i.
3). The use of external hold-down system is practical from the technological point of view. During the tests the force (F) and the relative slip of the components (d1. 2 Walls at the basis of the investigation: (a) fully sheathed. Experiments on wall-panel components 3. 2006 2440 2440 1170 2440 1940 2440 1200 3600 1200 3600 1200 160 1200 1200 1200 3600 1200 3600 1200 160 1200 160 1200 1200 a) b) Fig. which is fixed in the base concrete using anchor bolts and it is laterally connected to the wood panel with nails and/or woodscrews. Placing the hold-down profiles to both sides. (b) with door opening 3. with nails at 20 mm from the margin of the OSB. like in the symmetrical arrangement. while P-2 represents nailed connections in the central part of the OSB plates. 100 100 100 100 220 OSB 220 Stud 345 755 450 750 20 20 Nail 310 L profile 105 310 Stud 205 100 (a) P-1 100 100 Anchor (c) P-3 100 100 220 100 OSB 220 445 750 120 Nail 120 Stud L profile 210 105 100 100 Anchor (b) P-2 (d) P-4 Fig. 3 Tested wall-panel components The specimens were tested monotonically using displacement control. Both symmetrical (P-3) and asymmetrical (P-4) anchoring was tested. with loads intended to mimic the way components are loaded in the wall (Fig.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. P-2) represent the nailed connections between the OSB plate (t=15mm) and studs or tracks in the skeleton frame (80×130). The first sets of specimens (P-1. identical to the ones usually used in the wall panels. were used for the connections. Portugal November. 4). anchoring elements) as corresponding parts of the walls. The anchoring system is composed of an L profile (BMF-105). The second set of specimens (P-3. P-4) represents the typical hold-down detail of the walls. L=60mm). The subcomponents were made from the same base materials (wood.5mm. nails. as walls are finished on both sides from the workshop. OSB. Galvanised spiral nail (dnet=3. d2) was measured.1 Specimens and testing procedure Two types of components were considered to significantly influence the behaviour of the wall (Fig. 105 Stud 210 445 205 160 1200 1003 165 . Five specimens of each typology have been tested. is only possible at internal walls. P-1 represents nailed connections on the perimeter of the OSB plates.
7). The eccentricity between the bolt and the fixing to stud resulted in reduced rigidity and strength of the hold-down. and the elastic displacement corresponding to K and Fel. 4 Loading and measurements. Connection group P-2 Specimen P-2/1 P-2/2 P-2/3 P-2/4 P-2/5 Average K (N/mm) 511 976 951 682 831 817 Fel (N) 886 1190 915 959 963 988 Duct 11 14 15 11 14 14 Fmax (N) 985 1322 1016 1066 1070 1098 For the hold-down detail several configurations have been tested. d1 d2 d1 1400 1200 1000 F/2(N) F/2(N) 800 600 400 200 0 0 10 d(mm) 20 30 P1/1 P1/4 P1/2 P1/5 P1/3 1400 1200 1000 800 600 400 200 0 0 10 d(mm) 20 30 P2/1 P2/4 P2/2 P2/5 P2/3 a) b) Fig. The force-displacement diagrams are presented in Fig. The elastic strength (Fel) was considered to be 90% of the maximum capacity reached during loading (Fmax). Table 1.2 Results of the component tests d1 d2 Specimens in groups P-1 and P-2 behaved in essentially the same way. Table 2). The behaviour of the connections was very ductile. Connection group P-1 Specimen P-1/1 P-1/2 P-1/3 P-1/4 P-1/5 Average K (N/mm) 843 940 694 805 804 817 Fel (N) 787 968 968 1220 1020 993 Duct 20 18 16 15 11 16 Fmax (N) 875 1076 1075 1355 1134 1103 Table 2. In case of the first specimen in the groups P-3 (P-3/1) the L profile was fixed to the base with a single bolt in the hole near its margin (Fig.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. observed. the edge distance of the OSB did not influence the behaviour. 6). in the laboratory of the " Politehnica"University of Timisoara. 166 . but only in the final stage. 5 Characteristic curves for connections: (a) P-1 and (b) P-2 Conventional rigidity and strength have been calculated for the connections (Table 1. Because the failure of the connection was due to bending of the nail and cutting of the nail into the wood. Portugal November. F Exceptions were specimens P-1/2 and P-1/4 when F F cutting of the nail trough the edge of the OSB was Fig. P-3 and P-4 bending of the nails. 5. The rigidity (K) was calculated as the secant stiffness to the force level of 0. d2 3. In order to improve the behaviour. 2006 F F F The specimens were tested with a TESTWELL universal testing machine. with computer control of the experiment and of the measurements.6×Fmax. where the displacement was calculated as (d1+d2)/2 and the force as F/2 to reduce it to a single nail. The ductility of the connection (Duct) was determined as the ratio between the displacements at Fel in the unloading branch of the curve. in the subsequent specimens two bolts were used to connect the L profile to the base in the vicinity of the stud fixing (Fig. after extensive Specimens P-1.
P-4/4). 8 Characteristic curves for hold-downs: (a) group P-3 and (b) P-4 The procedure for calculating rigidity. 10000 10000 8000 8000 F/2(N) 6000 F(N) 6000 4000 4000 2000 2000 P3/1 P3/4 0 10 P3/2 P3/5 20 d(mm) P3/3 0 P4/1 P4/4 P4/2 P4/5 20 d(mm) P4/3 0 30 0 10 30 a) b) Fig. Properties of anchors in group P-4 Duct 4 4 6 6 4 5 Fmax (N) 3892 8443 9431 9139 5708 9004 Specimen K (N/mm) Fel (N) Duct Fmax (N) Specimen K (N/mm) Fel (N) P-3/1* 1288 4876 5 5418 P-4/1* 514 3503 P-3/2 2808 7156 11 7951 P-4/2 1074 7598 P-3/3 2283 6653 11 7393 P-4/3 1939 8488 P-3/4 2760 9056 8 10062 P-4/4 1939 8225 P-3/5* 1675 4935 7 5483 P-4/5* 1182 5137 Average° 2617 7622 10 8469 Average° 1651 8104 * Non-standard fixing of the L profile (see description in text). P-4/5) or the wood screw only (P-4/1). P-4/2. To analyse the effect of weakened connections tests were also made using six nails (P-3/5. Portugal November. 6 Deformation of P-3/1 due to Fig. 8. P-4/3. P-3/3. 1005 167 . Properties of anchors in group P-3 Table 4. 7 Improved behaviour of specimens from P-3/2 eccentricity Six spiral nails and a wood screw were used as standard fixing between the L profile and the stud (P3/2. Table 3. Characteristic curves of specimens in both groups are presented in Fig. P-3/4. ° For the calculation of the average values the cases of non-standard fixing were ignored. Resulting characteristic values are presented in Table 3 and Table 4.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 2006 Fig. strength and ductility was identical as for specimens in group P-1 and P-2.
see Table 1). The nailing interval was 200mm. The OSB panels can not deform as their inplane rigidity is very large. In wall-panel calculations this connection is often considered pinned.1 Preparation of the FE model Deformation pattern of the wall under lateral load is presented in Fig. In the real wall the connection of the studs to the tracks is made with nails which run along the grains of the stud. The contact of two consecutive OSB panels along the vertical edge was taken into account. Based on the observed deformations (Fig. =950kg/m3. the benefit being that FEM provides detailed information on the performance of various components. In the model (Fig. 2. in this figure and all subsequent figures presenting the FE model the inner layer of OSB is removed from the representation. 10) the frame was sheathed on both sides with OSB.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Only the case of hold-down on one side of the wall was modelled. instead they undergo rigid body rotation. The differential deformations of the sheathing and frame have to be accommodated by the nailing of the OSB to the frame. =480kg/m3 and OSB E=9000 N/mm2. The fasteners. which connect the frame and the OSB were modelled as axial connectors with the stiffness and resistance from the component-tests (K=817N/mm. in the FEM both properties have been concentrated in the connector elements between the L and the stud. Therefore. The hold-down of the walls was modelled in three ways: (i) full fixity of the base track. (ii) BMF-105 profiles at each stud and (iii) BMF-105 profiles at intermediate studs but stronger BMF-KR-285 fixings at end studs. 2006 4. 168 .a. assumption which is not realistic in this case. v y H v vH L vH Fig. 10 FE model with OSB panels from inner side removed from the representation The configuration of the wall corresponds to the one presented in Fig. The wood frame (80×130mm) was modelled using elastic hexahedrons (C3D8R). 9. The lateral load applied at the top causes the frame to distort to a parallelogram. a FE model to predict the behaviour of the fully sheathed wall panel was developed in ABAQUS 6. As reference case in the modelling the connection was considered pined but later it was modelled as flexible in the nails withdrawal direction along the stud' s axis (Table 5). This connection resists tension only by the withdrawal of the nails. and the OSB using homogeneous linear quadrilateral shell elements (S4R) with the thickness of 15mm.5 [6]. Fel=300N. but for clarity. The flexibility of the hold-down connection was evaluated based on the corner detail experiments described in Chapter 3. 9 Typical deformation of the wall Fig. The properties of the components for the FE model were: wood frame E=11300 N/mm2. Fel=990N. The withdrawal properties of the connection were estimated to be K=200N/mm. The behaviour of the wall panel can be predicted using FEM with a general purpose modelling tool. 7) it was assumed that both the strength limitation and the source of flexibility come from the connection of the L profile to the stud. Axial connectors were used to model fasteners. Three vertically placed OSB panels were used on both sides of the frame. Portugal November. but it resists compression due to the contact of the two elements.2. In this study. Finite element modelling 4.
Further reduction of the rigidity was observed due to the flexibility of the corner detail.b. Portugal November. when the withdrawal capacity of the stud to track connection was taken into account (W3p). Connection of BMF-105 to the track: K=2617N/mm. 60000 50000 40000 F(N) Elastic: E=9000 N/mm2 60000 50000 40000 F(N) 30000 20000 10000 0 30 30000 20000 10000 0 0 10 d(mm) 20 W3 W3p W4 W4p W4pp 0 10 d(mm) 20 30 a) b) Fig. 11. The same phenomenon was observed for the pair W4. and the capacity was limited to F=30kN (Fig.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 11. Fel=300N. ∞ Rigid connection of BMF-105 to the track. Fel=7622N.a). BMF-KR-285 to stud K=3×2617N/mm. The fundamental weakness of the hold-down in this configuration is that the BMF-105 profile is connected to the lower track only. BMF-105 to track: K=2617N/mm. the failure took place in the lower connector of the marginal stud on the uplifted side of the wall. Fel=2×7622N. Fel=7622N. 12. W4p (Fig.a). Fel=7622N. W5p and W5pp (Table 5). Name Frame OSB OSB nailing Elastic-Plastic: K=817N/mm. BMF-KR-285 to stud K=2×2617N/mm. but the capacity was not changed because the failure still took place in the connection between the stud and the lower track (Fig. The improved corner detail was used for W5. Alternatives of the FE model of the wall Hold-down to Vertical force Frame Fixing (KN) W3 Continuous W3p fixing W4 BMF-105 ∞ profile for each W4p stud W4pp Withdrawal* W4p-v 56 56 W4pp-v W5 Pinned ∞ BMF-105 profile for W5p middle studs + W5pp Withdrawal* long BMF .2 Results of the modelling In the reference case of base track fixed continuously to the base and studs connected to tracks with pinned connectors (W3) the failure took place in the connections between the OSB and the skeleton at approximately F=45kN (Fig. In order to improve the behaviour of the wall the use of higher hold-down BMF-KR-285 was proposed at the end studs where the uplift forces are the largest. Fel=990N Stud-Track Connection Pinned Withdrawal* Pinned Hold-down Elastic: E=11300 N/mm2 4. Fig. 11. Similar failure mode was observed in tests by Salenikovich [7]. BMF-105 to track K=2617N/mm.b). 11. For W4pp the hold-down corner was modelled to take into account the rigidity and capacity of the fixing between the BMF-105 profile and the track (Table 5). 2006 Table 5.KR W5pp+ 285 profile for W5p-v 56 end studs W5pp+-v 56 * In the direction of nail withdrawal: K=200N/mm. However. This failure mode is undesirable as it is localised and does not allow the full load bearing potential of the sheathing to develop.a). 11 Effect of limited withdrawal resistance of the stud to track connection 1007 169 . Fel=3×7622N.
Up to this point the effect of vertical force.b). In this case of W5pp+ the favourable mode of failure of the OSB to frame connections has been obtained (Fig. It can also be observed that the characteristic curves are very similar for W5. However. This shows that the failure is not taking place in the stud to track connection. was not taken into account. 14 Effect of vertical forces (V=56kN) It has to be observed that the increase of the capacity of the OSB to skeleton fixings would result in increased demands on the hold-down. As it can be seen in Fig. the effect of the vertical load is significant when originally the failure was in the hold-down detail ( W4pp vs. 60000 50000 40000 F(N) 30000 20000 10000 0 0 10 d(mm) 20 30 60000 50000 40000 F(N) W5 W5p W5pp W5pp+ 30000 20000 10000 0 0 10 d(mm) 20 30 W4pp W4pp_v W5pp+ W5pp+_v Fig. W5p (Fig. compared to the basic fixing of BMF-105. 12 Deformed shapes of (a) W4p and (b) W5pp+ at top horizontal displacement of 40mm The rigidity and capacity of the connection between the corner hold-down and the stud can be increased by increasing the number of nails and wood bolts. 14. In W5pp and W5pp+ it was supposed that the rigidity and capacity of the fixing with BMF-KR-285 is increased two or threefold respectively. W4pp_v). 2006 a) b) Fig. 170 .International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 13 Improvements by using BMF-KR-285 Fig. models corresponding to W4 and W5 were analysed with a total vertical force of 56kN distributed to the studs (Table 5). In order to account for this effect. 13). 12. and the corner detail should also be reconsidered. The possibility to fix the hold-down directly to the stud reduced the flexibility. W5pp_v). Portugal November. which contribute to the reduction of the uplift force in the corner hold-down. but has limited influence if originally the failure was taking place in the OSB to frame nailing (W5pp vs. possibly strengthened. the level of force in W5pp still does not reach the level of 45kN due to the failure of the connections between the hold-down and the stud. The number of screws and nails connecting the hold-down to the stud can be increased three times due to the shape of BMF-KR-285.
a). There is interaction between the loaded wall and all perpendicular walls intersecting it. 15) curves are presented in Fig. 2. 17. Portugal November. The rotation of the OSB plates was observed for each specimen (Fig. the walls followed the expected deformation pattern as presented in Fig. 9. 2006 5. This is more significant than the 5-10% in the case of steel skeleton walls [8] most probably because in the reverse cycles the bending of the nail takes place to some extent. in reality there are several secondary load mechanisms which may prevent sliding but they are neglected in the testing procedure. Hence. due to their eccentricity in the horizontal plane were not able to prevent the horizontal sliding of the wall on its support.1 Test results In the very early stage of testing it became clear that the flexibility of the hold-down connections was much larger that the FEM predicted. 17. H2 H1 H4 H3 V1 V2 Fig. Also. the dissipative capacity of the timber walls is slightly larger. These are: • • • Increased friction between the track and the concrete support. Fig. It was impossible to evaluate without further tests if these secondary load paths effectively prevent sliding in a real structure. External walls. with large windows are flexible and their contribution to the load bearing of the house is less significant. The lateral force (F) versus top horizontal displacement (d = (H1+H2)/2. 17. For the middle studs BMF-105 hold-down was used (Fig. The sliding of the wall (H3. Findings from the FE analysis have been incorporated in the detailing of the hold-down. 15). but such fixing is difficult to implement in the real structure. The testing was undertaken at the " Politehnica"University of Timisoara. while end studs and studs near the door opening were fixed with BMF-KR 285 fixing (Fig. 17. When sliding on the support was prevented. 16 Spec2 before test 5.c). In order to prevent the sliding centric fixing of the lower track had to be provided. The behaviour is characterised by strong pinching with the pinching force representing 15-20% from the maximum force. The tests were conducted in displacement control. the lateral load in LC and horizontal top displacements (H1. The fixings.b) and in the hold-down corner due to repeated cycles in case of both specimens (Fig. 15 Wall placed in the loading frame Fig.a). 17. Usually a 5 cm finishing concrete is poured inside the rooms which encases the track. The walls were fixed at the bottom and loaded at the top using the load cell LC (Fig. Instead of the expected uplift deformations. the hold-downs deformed excessively in shear.c).International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. A fully sheathed wall (Spec1) and a typical interior wall with door opening (Spec2) were tested. 171 1009 . H4) and the uplift of the corners (V1. V2) were also measured. 18. Supplementary deformations strongly concentrated in the link area for Spec2 (Fig. Full scale wall-panel tests In order to validate the findings of the FE analysis two full scale tests have been performed on the wall panels presented in Fig. H2) being recorded.
The FE models highlighted several the hold-down due to repeated cycles weaknesses of the originally proposed corner detail. The model also predicts the end of the elastic behaviour. 2006 a) The positive and negative envelope of Spec1 can be compared in Fig. (c) damage of even if FE results still need to be viewed with caution. The finite element modelling proved to be a useful c) tool for predicting the behaviour of a wall panel Fig. 5). (b) link area deformation. b) For the studied sizes the nailed connection failed by bending of the nail and cutting of the nail in the fibbers of the wood. If large eccentricities are present. 6. The used FE model was identical to W5PP+. the hold-down is weakened. 19 to modelling results.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 90000 60000 30000 60000 30000 F(N) 0 -25 -15 -5 -30000 -30000 F(N) 0 -50 -30 -10 10 30 50 5 15 25 -60000 -90000 -60000 a) d(mm) b) d(mm) Fig. This is the preferred failure mode because edge failure of the OSB would provide limited ductility leading to increased seismic vulnerability of the structure. This overstrength is related to the supplementary capacity of the individual nails over the characteristic strength value of 817N used in the modelling (Fig. The tests on the hold-downs emphasized the need to avoid eccentric load transmission paths. but the sliding of the lower track was prevented. 17 Observed damage: (a) sliding of the based on the characteristics of the components. but does not account for the significant non-linear overstrength. 18 Hysteretic curves for the two wall configurations: (a) Spec1 and (b) Spec2 172 . Conclusions The experimental results on nailed OSB to frame connections and on hold-down details showed good homogeneity and ductile behaviour. OSB. Apparently. These details can be used in earthquake situation. Portugal November. the prediction of the rigidity is satisfactory but the FE model does not take into account the initial stiff portion of the response of the wall.
PhD thesis Faculty of the Virginia Polytechnic Institute and State University. Zacher E. Based on the studies the following recommendations were formulated concerning the design: 90000 • • • • • • 7. Folz B.. CUREE-Caltech Woodframe Project Tissell J. OSB or wood quality is changed. Filiatrault A. On the other hand. University of California Forest Products Laboratory. Folz B.C. 1991 Foliente G. CUREE-Caltech Woodframe Project Fischer D. W-06 Shake Table Tests of a TwoStory Woodframe House. UK. Filiatrault A. rather high. For this particular configuration of OSB thickness.. Use BMF-KR-285 hold-downs at marginal studs and on both sides of door openings. nail type and wood quality. 19 FEM & experimental envelope curve. Volume 42. Portugal November. p. American Plywood Association: Tacoma. Verify by test if nails.5.. W-05 Two-Story Single Family House Shake Table Test Data. Seible F. February 2004..M.).. 1993 Wolfe R.W. (ed. 2006 An important conclusion of the FE analysis was that even an external anchorage is capable to assure sufficient uplift strength to force failure of the sheathing to frame connections. 1994 ABAQUS 6. CA. Washington.. level of detailing of the model. Wood Structural Panel Shear Walls.. XXII-1-19. When configuring the structure always transmit vertical forces to load bearing shear walls.J. VA. the 60000 question of sliding of the wall has not been answered satisfyingly by this study. it is not advantageous to increase the number of sheath to frame nails 30000 because the failure of the hold-down will govern Spec2+ Spec2the behaviour. References [1] [2] [3] [4] Fischer D.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Connect the BMF-KR-285 profile to the vertical stud with at least 3 wood screws. However. Part I: Experimental Research. APA Research report 154.. 20mm distance from the margin was sufficient. 1011 .5.C. Walford... Seible F. At this stage provide fixing to avoid sliding. 2004 Salenikovich A. Analysis Users Manual: Abaqus Version 6.. Minimise the eccentricities in hold-down connections by placing the anchors close to the track. Thin Walled Structures. 321-338 173 [5] [6] [7] [8] F(N) Ensure sufficient distance of the nails form the margin of the OSB to avoid brittle failure.. Proceedings of Research Needs Workshop: Analysis. Uang C...M. 2000 Fülöp L.C. Richmond. Design and Testing of Timber Structures under Seismic Loads. Performance of Wall-Stud Cold-Formed Shear Panels under Monotonic and Cyclic Loading.A. In Foliente G.. leading to reduced ductility.. Proceedings of Workshop on Full-Scale Behavior of Wood-Framed Buildings in Earthquakes and High Winds. Blacksburg. Uang C. Spec1+ Spec1FEM The test results on wall panels highlighted the 0 need to take into account in the FEM the non 0 10 20 30 linear overstrength of the components.. North American Structural Performance Tests of Low-Rise WoodFramed Building Systems. results is not satisfactory even at this. Performance Tests of Timber Structural System under Seismic Loads.G. Moody R.R. Dubina D. The d(mm) quantitative correlation between the test and FEM Fig.. The Racking Performance of Light-Framed Shear Walls. Number 2. Further studies to verify the sliding of the wall is necessary.
MDF>PB>OSB>Plywood for ultimate limit. The ranking order of material with regard to capacity was CBPB>PB>MDF>OSB>Plywood for stiffness. Portugal November. 1 Introduction Testing carried out using the EN performance method for floors (EN 1195) has shown that the performance of a floor subjected to concentrated loads and dynamic impact loads is affected by the efficiency of the jointing method between the boards. there was no significant relation between point load and impact load performance. However. The objective of this study was to thoroughly evaluate dynamic impact and static point load performance of various wood based panels. However. indicating that the most critical test location should be at the joints. This study detailed test methods to evaluate the dynamic performance (impact load) of various wood based panels and joint profiles. Vahik Enjily BRE International Unit. University of Bath. to examine the effect of different joint profiles on both static and dynamic impact load performance and to examine the correlation between dynamic impact and static point load performance of wood based panels. WD25 9XX Abstract The dynamic performance of wood based panel as a floor or roof can be critical with respect to the permissible deformation characteristics in design. BA2 7AY Dr. Two small-scale test methods (static point load and dynamic impact load) were developed and assessed. MDF>PB>OSB>Plywood>CBPB for serviceability.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Watford. The results presented in this paper are part 175 . 2006 Dynamic Performance of Wood Based Panels Dr Mizi Fan BRE Centre for Innovative Construction Materials. Building Research Establishment. it has become common practice to use static characteristic values for design. The results showed that the best performance capacity was very different between the types of panel products and joint profiles. This joint is most commonly the tongue and grooved (t&g) types. since this property has not been fully investigated and no standard test procedure has been developed. Panels without joints had a higher performance capacity than jointed panels. which vary in profile. The joint profiles 1 and 2 performed better than profile 4.
Figure 1: Set-up of static point load and impact load tests Static point load The test piece was placed horizontally and clamped rigidly along four edges. The load (F) was applied at a constant rate of deflection and adjusted so that the maximum load (Fmax) was reached in (300±120) seconds. The principle of the static load test firstly involves two cycles of a low level loading (Figure 2) which were applied to determine the stiffness of the test piece. 2 Materials and methods 2. Portugal November. 14. 2006 of outcomes from a complex EU co-funded project on the performance of wood based panels in construction. the appropriate static load was applied to the centre of the test piece and the corresponding deformations and damage were observed.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. The loading procedure was taken as the displacement of the loading disc relative to the bearing (rigid frame) at points 01. Testing was applied to both solid and jointed test pieces in order to determine the efficiency of the joint profiles used.1 Test methods Testing was carried out with the principle of the EN performance method for floors (EN 1195) subjected to concentrated loads and impact loads (Figure 1). 11. The serviceability limit (Fser) is reached during loading if a 176 . Following this (after point 24) a gradually increasing load was applied until the serviceability limit and failure (ultimate limit) occurred. 04. 21 and 24.
The falling body was raised 25 mm from the surface of the test piece and allowed to drop freely. Figure 2: Loading procedure for stiffness test. Figure 3: Loading procedure for maximum and serviceability limit loads Hard body impact load A test piece was rigidly clamped along its edges and a body of a defined mass and shape was dropped from a series of increasing height onto the centre point of test pieces until the test pieces was penetrated by the body or until the maximum drop height of 900 mm was reached without any significant damage occurring. and the head of the falling body impacted the test piece once 177 .International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. Portugal November. 2006 discontinuity (indication of irreversible damage) occurs in the load deflection curve even if there is no visible damage (Figure 3).
An initial deflection was measured before any impact on the test pieces and deflection was recorded every 100mm increase in drop height thereafter. resulting in a total of 500 tests. Portugal November. 2. This operation was repeated by increasing the drop high by 25 mm on each impact. 2006 per operation. Table 1: Table Panel products used in the project Material Abbreviation EN product Thickness Density (mm) 18 18 300 18 (kgm-3) 790-800 720-740 620-640 standard Particle Board PB EN 312-5 EN 622-5 EN OSB/3 Softwood Plywood PLY EN 636-3 EN634 Medium Density Fibreboard MDF Orientated Strand Board OSB 18 18 550-600 1200 Cement bonded particleboard CBPB 178 . A RaytestPROFILE was used to measure the density profile by monitoring the intensity of the radiation travelling through the test pieces. All of the test panels were conditioned at (20±2)°C/(65±5)% relative humidity until a constant weight was achieved.2 Materials Five commercially produced panel products related to the European product standards for wood-based panels were tested (Table 1). Density profile Density profile measurements were determined by cutting test pieces of 50x50 mm2 from the test pieces after impact or static point load tests. Each of these materials was machined with four types of T & G jointing profile and one solid. until failure of the test piece or reaching the maximum 900 mm drop height limit. with 10 tests being conducted on the tongue side of the joint and 10 on the groove side.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. with a total of twenty replicates for each configuration. The application of impact load was applied to each side of the joint.
i. There are two density values relevant to the strength: average values and density profiles. D=16 mm OSB and E=22 mm PB. i. A homogeneous wood based panel material. 2006 The principle of the tongue and groove profiles used is shown in Figure 4. Portugal November. The jointed test pieces were glued together with a standard PVA adhesive and used in accordance with the manufacturer instructions. A=16 mm PLY. see Figure 5 for the materials tested.1 Density profiles Density is an important physical characteristic of wood based panels. This means that they may perform better than other materials tested. B=18 mm CBPB.e. Figure 4: T&G profiles of test pieces 3 Results and Discussion 3. average value and density profile is similar. In Figure 5. performs better for force transformation in comparison with a panel with low density in the centre. because the tongue is manufactured in the weakest part of the panel across thickness. It is apparent that the density of MDF and plywood are more uniform throughout the thickness of the panels.e. C= 22 mm MDF.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 179 . average value is higher than the density profile in the middle of the panel.
International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. 2006 A B C D E Figure 5: Density profiles of materials tested 180 . Portugal November.
It can be see that there are significant differences between the stiffness of the panel products tested. The difference of stiffness between glued and unglued PB and CBPB were more significant than other materials tested. 21 and 24. W04. The ratio of the solid and jointed test pieces showed that there is 0 . F24 is the load (in the strength test) of 0. and W24 are the deformations recorded at the points 01. W21.3 Fser Serviceability load Average serviceability load are summarised in Table 3. The ranking order for the serviceability load is: MDF (highest)>PB>OSB>Plywood>CBPB. W24 and W21 are the deformations determined at loads F24 and F21 respectively. The reduction is highest for plywood. 181 .4 Fmax and F21 was the load (in the strength test) of 0. In general. The reduction is dependent on the types of materials and t&g profiles. 2 and 3 were significantly better than t&g 4 for both glued and unglued joints. It can be seen that there are significant differences between serviceability load limit of the panel products tested. [ ] 3.2 Stiffness and deformation The deformation at the load F = 0. W01. The solid test pieces are generally the stiffest and the glued profiles perform better than unglued.40 % reduction for glued and 0% . 04. The solid test pieces for all the materials showed higher serviceability than the jointed test samples. t&g 1. 2006 3. The stiffness was calculated from the equation: ⎧ ⎪ F24 − F21 ⎫ ⎪ Rm = ⎨ ⎬[ N / mm] ⎪ ⎩ W24 − W21 ⎪ ⎭ where.1 Fmax. t&g 2 (glued) seems to be the best for PB and CBPB. The various profiles performed differently on the materials tested. Mean test values are summarised in Table 2. Portugal November.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.4 Fmax was calculated as the average of the first and second loadings as follows: ⎧ W04 −W 01 ] +[W24 − W21 4⎪ Wm = ⎨ 3⎪ 2 ⎩ [ ]⎫ ⎪ ⎬[ mm] ⎪ ⎭ where. except for OSB.60% for unglued. The ranking order of stiffness is: CBPB (stiffest)>PB>MDF>OSB>PLY.
2004 glued t&g1 t&g2 t&g3 t&g4 1802 1866 1778 1515 0.13 540 0.22 Table 3: Serviceability load Joint Mean (N) Cov (%) Mean (N) Cov (%) Mean (N) Cov (%) Mean (N) Cov (%) Mean (N) Cov (%) 22mm PB Solid 8277 0.11 0.28 182 .10 1964 1874 1689 1611 0.18 2144 0.25 0.14 t&g1.11 0.6965 glued t&g2.2167 glued t&g2.05 522 0.7269 glued t&g3.31 2117 0.16 878 0.19 t&g1.06 2309 0.19 0.30 0.14 16mm OSB 3541 3693 0.27 0.14 0. 2006 the glued profiles performed better than unglued.19 4040 0.12 0.08 0.13 16mm OSB 935 889 0.27 0.16 0.13 758 735 769 711 0.18 6840 0. Table 2: Stiffness of materials tested Joint Mean (N) Cov (%) Mean (N) Cov (%) Mean (N) Cov (%) Mean (N) Cov (%) Mean (N) Cov (%) 22mm PB Solid 2560 0.06 3847 0.18 0.07 0.13 22mm MDF 9065 6987 0.15 3885 0.16 2089 0.11 586 0.11 15mm PW 864 593 0.14 18mm CBPB 2779 2738 0.15 0.11 1486 0. t&g 1 and 2 seem to be the best on all materials.15 0. 2 and 3 are significantly better than t&g 4 for all materials except for Plywood. whether glued or unglued.23 0.11 545 495 485 356 0.13 0.09 15mm PW 4515 3422 0.10 807 0.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.20 0.2231 glued t&g4.6900 0.13 0.32 0.12 18mm CBPB 4384 3999 0. The various profiles performed differently on the materials tested.17 0.20 0.11 4105 0. t&g 1.28 22mm MDF 2250 1890 0.13 0.15 2844 0.30 861 0. Portugal November.12 0.05 0.09 0.16 0.07 0.17 3817 0.09 6439 0.2110 glued t&g3.18 1689 1908 1606 1384 0.17 0.10 3723 0.
6271 glued t&g1 t&g2 t&g3 t&g4 6288 6177 5728 5851 0.11 0. The reduction of the 5% Fser and 5% Fult is almost identical.06 4670 0.6920 glued t&g1 t&g2 6548 6719 0. Table 4: Ultimate load Joint Mean (N) Cov (%) Mean (N) Cov (%) Mean (N) Cov (%) Mean (N) Cov (%) Mean (N) Cov (%) 22mm PB Solid 8408 0.26 0.08 3073 3295 2666 2817 0. The solid test pieces showed higher ultimate limit than the jointed test pieces.07 0.05 3184 3121 2813 2743 0.13 3453 0.12 15mm PW 5937 4409 0. t&g1 and 2 (glued) seem to be the best for all materials tested.13 0.09 0.7819 glued t&g3.15 4117 0.10 22mm MDF 9135 7136 0.7489 glued t&g2. Fult is in general higher than Fser.13 0.17 0.14 3558 3328 0.13 0. The various profiles performed very similarly for the materials tested.07 5314 0.12 3703 0.13 4161 0.08 3334 3320 0.08 5734 0. the glued profiles performed better than the unglued.19 3.19 0.09 18mm CBPB 4410 4324 0.23 0.19 0.15 3686 0.15 0.22 0.18 0. In general.14 16mm OSB 4155 3987 0. 2006 glued t&g4.06 0.16 0. Portugal November.05 0.16 4121 0. The ranking order is: MDF (highest)>PB>OSB>Plywood>CBPB.23 0.7343 glued t&g4.14 0.09 4562 0.18 0.18 4043 0.13 0.4 Fult Ultimate load There are significant differences between the ultimate limit load (Table 4).09 3456 3637 0.12 6175 5890 0.12 3194 0.10 183 . 2 (and 3) are significantly better than t&g 4 for all materials tested.13 0.07 0.10 4183 0.07 0.11 3054 2959 2368 2442 0.13 t&g1.11 6587 0.19 0.12 0.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.14 3635 0.15 6961 0.17 5675 5755 5145 4370 0.15 0. However.16 0.08 0. t&g 1.
05 3040 3198 0. The unglued profile 2 performed better than other profiles for OSB. The level of performance with the OSB was relatively consistent across all of the configurations. PB. each of the materials achieved best maximum drop height with a unique combination of joint profile and/or use of adhesive in the t&g joint. 184 . The solid plywood performed better by an average of 30% than the jointed (glued/unglued) test panels. 3. Portugal November. With the exception of CBPB. T is the nominal thickness of the test piece (mm). The performance of the solid panels was the lowest for MDF and OSB.12 0.16 3. the use of adhesive in the t&g joint had an adverse effect on the impact performance.5 Dynamic impact strength The hard body impact test results are expressed as a measure of impact strength. and highest for Plywood and CBPB. MDF and Plywood performed best with the unglued profiles 3 and 4 respectively. with the exception of CBPB.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.10 5434 4549 0. Hmax is the drop height (mm) required to produce failure of the test piece. but reduces impact strength by up to 30% for other materials tested.07 0.12 0. Overall ranking order for the hard body impact strength performance (best performing first) is: Plywood>MDF>PB>OSB>CBPB. This is due to the fact that increasing stiffness of the test piece that occurs from applying the adhesive reduces the capacity of the test piece to absorb the impact energy. Each material achieved best results with a unique combination of profile and/or use of adhesive in the t&g joint. The hard body impact results from the program of testing showed that there is a significant difference in the impact strength between the panel products tested.07 2875 3132 0.6 Maximum drop height As for the strength. The introduction of adhesive into the joint had a largely negative effect on the drop height achieved before failure. 2006 t&g3 t&g4 5976 6420 0.16 0. gluing t&g joint can improve the performance by approximately 15% for CBPB and PB.06 3499 3118 0. this is calculated from the equation: f = H max T Where. Profile 1 gave average results for all materials.09 0. Overall.
7 Failure modes Examples of failure modes are given in Figure 6. Impacts caused an indentation of the surface of the plywood. with the surface of the panel suffering relatively minor indentation damage. For OSB the panel failed due to the damage of area of the impact which suffered severe indentation damage at higher impact heights. Plywood panels suffered minor indentation damage. further impacts caused cracking around the area of impact and delamination of core material leading to failure of the panel. For PB and MDF. followed by cracking/crushing of the core material and failure of either the tongue or grooved component. then panels fail and ultimately were smashed into pieces. followed by cracking surrounding the area of impact which led to panel/joint failure. further impacts caused failure of the joint which fractured upon impact. For PB. For OSB initial impacts caused indentation. followed by localized deflection causing the plywood surrounding the plastic component to delaminate and ultimately crack with subsequent impacts. CBPB suffered the same surface damage as with profiles 1 and 2. not the joint. 2006 3. with failure of the panel occurring from large indentations built up by successive impacts. In the worst cases the impacts caused the panel to break up. Impacts caused the plastic component of the joint of CBPB panels to crack. causing localized crushing of the board. 185 . Plywood suffered damage to the localized area of the panel that deflected under higher impact loads to a point where the joint would slip and the impacts produce a localized joint failure. MDF and OSB this profile remained intact.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. CBPB firstly cracked the upper surface layer near the area of impact and further impacts caused damage to the tongue together with deeper cracks. For t&g profile 1 and 2. eventually leading to failure of the material. Portugal November. For t&g 4. spreading along the surface of the panel. with failure occurring through punching shear. Initial impacts dented the surface of PB and MDF. For t&g profile 3. the impact energy caused the plastic component to crack and fail.
186 . Portugal November. 2006 Figure 6: Example of failure modes under impact load 3.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra.8 Comparison of static point load with impact load Serviceability The overall results are given as the relationship 1) between the mean deflection at the serviceability limit for static point load and the drop height in mm required for a 1 mm deformation for hard body impact. 2) between the drop height in mm required for a 1 mm deformation for impact load and the type of panel product (Figure 7).
It can also be seen that there is no correlation between the impact under 1 mm permanent deformation and panel types. ult for impact load and PLY16 1 2 3 4 5 6 7 8 187 . the deflection under the static point load is elastic and the 1 mm deformation under impact load is permanent.and impact load. It can also be seen that there is no correlation between the ratio of Fult for static point load to H panel types tested. The Figure shows that there is no significant correlation between the deflection under static point load and the impact height for a permanent 1 mm deflection under hard body impact load.International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. It can be seen that there is no significant correlation between the ultimate limit properties between static point. Ultimate limit Figure 8 show relationships between the mean ultimate load (Fult ) for static point load and the mean ultimate drop high (H ult) for hard body impact load. Portugal November. 2006 Serviceability limit IMPACT LOAD mm drop heigh for 1 mm deformation IMPACT LOAD mm drop heigh for 1 mm deformation 1200 MDF22 OSB16 tg4 tg4 Serviceability limit 1200 solid tg1-glued tg2-glued tg3-glued tg4-glued tg1 tg2 tg3 tg4 PB18MDF22 MDF22 PB18 PB22 PB22 MDF22 PB22 PB22 MDF22 PB22 PB18 PB22 PB18 PB22 MDF22 PB22 PB18 PB18 PB18 MDF22 PB18 PB18 OSB16 MDF22 OSB16 PB22 MDF18 OSB16 MDF18 MDF22 OSB16 OSB16 OSB16 OSB16 OSB16 MDF18 MDF18 MDF18 PLY16 MDF18 MDF18 MDF18 MDF18 PLY16 1000 1000 tg1tg4-glued PLY16 PLY16 800 600 CBPB18 CBPB18 CBPB18 CBPB18 CBPB18 CBPB18 CBPB18 CBPB18 CBPB18 tg4 tg1-glued tg4-glued tg3 solid tg1 tg3-glued tg2-glued tg2-glued tg2 tg1-glued tg3-glued tg1-glued tg3-glued tg3 tg2-glued tg1-glued tg4-glued tg1 tg3 tg4 solid tg4-glued tg2 tg1 tg3 tg1-glued tg2 tg4-glued solid tg2-glued tg3-glued tg3 solid tg1 tg4 tg4-glued tg1-glued tg1 tg3 tg3-glued tg2-glued tg2 PLY16 PLY16 PLY16 solid tg2 solid PLY16 tg2 800 tg4 tg2-glued tg3-glued PLY16 600 400 tg4 tg1-glued tg3-glued solid tg2-glued tg4-glued tg2 tg1 tg3 400 200 200 CBPB18 PB18 PB22 MDF18 MDF22 OSB16 STATIC POINT LOAD mm deflection at Fser Figure 7: Serviceability limit between impact and static load In Figure 7.
International Workshop on "Earthquake Engineering on Timber Structures" Coimbra. • Panels without joints have a higher performance capacity than jointed panels. therefore the most critical test location is at the joints. MDF>PB>OSB>Plywood>CBPB for serviceability. • It will be possible to design a t&g profile for wood-based panels for structural use supporting the Construction Product Directive (CPD). 188 MDF22 PLY16 solid tg1-glued tg2-glued tg3-glued tg4-glued tg1 tg2 tg3 tg4 . • There is no significant relation between point load test and impact test. • Ranking of material with regard to capacity is CBPB>PB>MDF>OSB>Plywood for stiffness. Portugal November. MDF>PB>OSB>Plywood for Ultimate limit. 2006 Ultimate limit 1200 MDF22 Ultimate limit 12 POINT LOAD/IMPACT LOAD Fult/Hult tg4 IMPACT LOAD Hult Maximum drop heigh in mm PLY16 PLY16 PLY16 PLY16 1000 900 800 700 600 500 CBPB18 MDF22 MDF18 MDF18 MDF22 PB18 MDF22 PB18 PLY16 PB18 PB22 MDF22 PB18MDF18MDF18 MDF18 PB22 MDF18 PB22 MDF22 PB22 MDF18 PB18 PB22 MDF22 1100 11 10 9 8 7 6 5 4 3 PB18 PB22 CBPB18 PLY16 PLY16 PLY16 PB18 PB18 PB18 tg4 tg1 tg2 OSB16 PLY16 OSB16 OSB16 OSB16 OSB16 tg1-glued tg1 tg2 PB18 tg3 tg2 tg4 tg3 tg4 CBPB18 CBPB18 CBPB18 tg1 tg4-glued tg1-glued tg2-glued tg3-glued solid solid OSB16 OSB16 CBPB18OSB16 CBPB18 CBPB18 CBPB18 tg4 tg4-glued tg2 tg1 tg1-glued solid tg2 tg1 tg1-glued tg2-glued tg1-glued tg3-glued tg2-glued tg3-glued tg1-glued tg4-glued tg3-glued tg2-glued solid tg4 tg4-glued solid tg2-glued tg3-glued solid PB22 MDF18 PB22 tg4-glued tg3 tg3 tg3-glued tg2-glued MDF22 MDF18 PB22 PB22 tg3 MDF22 tg2 tg4-glued tg3tg1 solid 400 300 200 100 0 2000 tg2-glued tg3-glued tg1-glued tg2 tg3 tg4tg1 tg4-glued CBPB18 OSB16 MDF18 3000 4000 5000 6000 7000 8000 9000 10000 STATIC POINT LOAD F ult Maximum load in N Figure 8: Relationship between maximum impact drop height and static point load 3 Conclusions • It can be concluded that the best performance capacity for static impact point load independence of panel product type is found for the profiles 1 and 2 and the worst for profile 4.
desenvolve ainda softwares à medida e dispõe de softwares vocacionados para várias áreas: Gestão escolar – ABC Gest e mobiliário de cozinhas – Promob. Viana do Castelo.madeicavado. Actualmente detém quatro empresas. Para mais informações. nomeadamente estruturas em madeira e revestimentos de fachadas ventiladas. a Damadeira. Cada uma das empresas está especializada numa área específica. A Tecniwood é a empresa do grupo mais vocacionada para a construção. quer em madeira maciça. Operam em vários pontos do país – Braga. Para além de deter produtos próprios. Indústria das cozinhas. a Timbermade e a Desis compondo o Grupo Madeicávado. O grupo conta ainda com a Desis.pt . orlagem.Grupo Madeicávado Madeiras A Madeicávado está no mercado há 20 anos no sector das madeiras e derivados. visite www. revestimentos. prestando serviços ao nível da maquinação da madeira. Neste momento está a expandir-se a novas áreas – acústica. corte. fresagem e secagem das madeiras. como pavimentos. A Timbermade constitui a única unidade fabril do grupo. carpintaria. Póvoa de Varzim e Alenquer. mobiliário em HPL. todas elas complementares umas às outras. construção civil. A Damadeira está especializada no comércio de madeiras e derivados e actua em vários sectores onde a madeira é protagonista: indústria do mobiliário. a Tecniwood. reabilitação e bio construção. perfuração. componentes de carpintaria e de construção. quer em HPL. uma empresa de informática que para além de assegurar esta componente no grupo. ainda serve o grupo. arquitectura e decoração.
utilizar de modo eco-eficiente a madeira virgem. Com 34 unidades industriais e uma capacidade instalada de 9. • Aglomerado de fibras duro (hardboard).6 milhões de m3 por ano. decoração. Suiça e África do Sul. Esses são os princípios que orientam a Sonae Indústria para a gestão e uso sustentável dos recursos naturais. espalhados por 10 países: Portugal. França. nomeadamente: • aglomerado de partículas de madeira (particleboard). Alemanha.componentes. As preocupações ambientais e o desenvolvimento sustentável fazem parte do compromisso da Sonae Indústria e são uma realidade na gestão da sua actividade. Canadá. a presença internacional e a experiência global da Sonae Indústria aprofundaram o seu conhecimento dos requisitos. Gerir sustentadamente a floresta.A SONAE INDÚSTRIA é um dos maiores produtores mundiais do sector dos painéis derivados de madeira. disponíveis no tempo e no local certos. Em 2005 o volume de negócios consolidado da SONAE INDÚSTRIA totalizou 1465 milhões de euros: No que se refere às vendas por tipo de produtos. a SONAE INDÚSTRIA empregava cerca de 6 mil colaboradores. A sua gama de produtos abrange. em todo o mundo.para as indústrias de mobiliário. • produtos e serviços de valor acrescentado . e ainda: • • laminados decorativos de alta pressão. tendências e padrões dos mercados e facilitaram o desenvolvimento dos produtos adequados. A 31 de Dezembro de 2005. Brasil. Espanha. A sua estratégia de negócio baseia-se portanto numa gama alargada de produtos e soluções para mobiliário. produtos químicos (formaldeído. decoração e para o sector de bricolage. resinas à base de formaldeído e papéis impregnados). • MDF (Medium Density Fibreboard). a distribuição em 2005 foi a seguinte: . • OSB (Oriented Strand Board). reutilizar e reciclar os materiais à base de madeira e valorizar como combustível os materiais de madeira não recicláveis. Reino Unido. construção. construção e embalagem e num excelente serviço aos clientes nestes segmentos de mercado. Holanda. soluções e sistemas .
o Magnum Board. Informação adicional pode ser obtida no n/ site www. o sistema de vigas I. A sua especialização na distribuição e transformação de pinhos escandinavos e russos. permite-lhe responder com grande rapidez às mais exigentes solicitações dos projectistas e profissionais.de elevada qualidade para fins estruturais. Pioneira na introdução no mercado ibérico de Engeneered Wood Products ( EWP ). o grupo Krono e a Simpson Strong-Tie. a partir dos seus 5 armazéns. o Thermowood. como a Finnforest. O seu Know-how é completado por parcerias estratégicas com empresas líderes mundiais nesta área. etc. bem como um serviço de logística próprio. dotada dos meios humanos e técnicos necessários para executar qualquer projecto mais exigente. o Kerto – LVL.JULAR . que cobrem todo o território nacional. o Glulam. com uma elevada componente de ID. o Douglas. a Jular inclui na sua gama produtos sofisticados como o OSB.35 anos a dar estrutura às suas ideias Fundada em 1973. a Jular tem na sua história uma forte tradição de utilização de pinhos – casquinhas .pt .jular. a Merk. A Jular dispõe na Azambuja de uma unidade de transformação. A Jular oferece um serviço de aconselhamento técnico. os ligadores Simpson.
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